Seismic Design of Steel Structures-complete course.pdf

Seismic Design of Steel Structures
Amit H. Varma and Judy Liu
CE697R
Fall 2012
MWF 2:30 – 3:20 PM
CIVL 2123
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Course Introduction
• Syllabus, Course Organization
• CE 697R Topics
• Introduction
• Basic Principles
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Syllabus Review syllabus; make sure that you
understand all course policies (e.g.
grading, ethics, etc.) and procedures in
event of an emergency.
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https://engineering.purdue.edu/Intranet/Groups/Administration/RPM/Safety/Classroom
EmergencyPlanning/CIVL @alirezasalehin
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Required Book
• Bruneau, M., Uang, C., Sabelli, R. Ductile
Design of Steel Structures, McGraw-Hill,
New York, NY, 2011.
http://www.michelbruneau.com/Ductile
%20Design%202nd%20Ed%20-
%20Errata.pdf
Errata (8/8/12 file also
posted to CE697 Dropbox)
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Additional References
• Will be made available in shared folder
on Dropbox or otherwise
Respond to e-mail with:
The e-mail address
associated with your
existing Dropbox account.
OR
E-mail address you’d like for
us to use in our invitation
to join Dropbox and shared
folder.
Please wait for
e-mail
invitation to
join Dropbox !!
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Some References
• 2010 Seismic Provisions for Structural Steel
Buildings, ANSI/AISC 341-10
• 2010 Prequalified Connections for Special and
Intermediate Steel Moment Frames for Seismic
Applications, with 2011 Supp. No. 1,ANSI/AISC 358-
10, with ANSI/AISC 358s1-11
• Seismic Rehabilitation of Existing Buildings,
ASCE/SEI 41-06
• NEHRP Recommended Provisions for Seismic
Regulations for New Buildings and Other Structures,
FEMA 450, 2003
• Recommended Seismic Design Criteria for New Steel
Moment-Frame Buildings, FEMA 350, 2000
• Minimum Design Loads for Buildings and Other
Structures, ASCE 7-10
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Course Project
Will also send e-mail requesting
information to help us form teams.
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Homework / Reading Assignments
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Files
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CE697R Topics
• Introduction and Basic Principles
• Structural Steel, Properties, Plastic
Behavior
• Moment Resisting Frames
• Steel Plate Shear Walls
• Braced Frames
– Concentrically, Eccentrically Braced;
Buckling-Restrained
• Analysis for Performance Evaluation
• Special Topics / Innovative Systems
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Acknowledgments
• Michael D. Engelhardt , Ph.D.
– Professor, University of Texas at Austin
– Eccentrically Braced Frames, with Egor
Popov, U.C. Berkeley
– T.R. Higgins Award for “Design of
Reduced Beam Section Moment
Connections.”
• AISC Educator Career Enhancement
Award to develop Teaching Modules
on Design of Seismic-Resistant Steel
Buildings
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Design of Seismic-
Resistant Steel
Building Structures
Prepared by:
Michael D. Engelhardt
University of Texas at Austin
with the support of the
American Institute of Steel Construction.
Version 1 - March 2007
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Introduction and Basic Principles
• Performance of Steel Buildings in Past
Earthquakes
• Codes for Seismic Resistant Steel Buildings
• Building Code Philosophy
• Overview of AISC Seismic Provisions
• AISC Seismic – General Requirements
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Earthquakes
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Causes of Earthquake Fatalities: 1900 to 1990
EERI slide series entitled: "Structural and Nonstructural Failures in Past Earthquakes."
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Recent Earthquakes
• 2010 Haiti Earthquake
• 2010 Maule, Chile Earthquake
• 2010 -2011 Christchurch, New Zealand
• 2011 Tohoku, Japan
– Steel Reinforced
Concrete (SRC) buildings
- Tsunami damage
industrial steel buildings
and residences
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http://www.aisc.org/uploadedcontent/2012
NASCCSessions/N9-1/
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Recent Earthquakes
• 2010 -2011 Christchurch, New Zealand
– 6 damaging earthquakes
– Steel structures generally performed well
– Most steel buildings constructed from
1990s (modern seismic codes)
– A few EBF link fractures, CBF brace fracture
(design/as-built detailing issues?)
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Fractured EBF links Intact gusset plate and endplate
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Why the good track record for steel?
• Little loss of life attributed to collapse of
steel buildings in earthquakes
• Likely causes? Steel structures …
– are generally lighter than masonry or RC. Lower
weight translates to lower seismic forces.
– typically show good ductility, even when not
specifically designed or detailed for seismic
resistance.
– have not been exposed as much to strong
earthquakes. Highly destructive earthquakes
around the world have generally occurred in areas
where there are very few steel structures.
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However ….
… modern welded
steel buildings had
shown an increasing
number of problems in
„recent‟ earthquakes.
Pino Suarez Complex
1985 Mexico City Earthquake
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1994 Northridge Earthquake
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1995 Hyogoken-Nanbu (Kobe) Earthquake
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•Approximately 90 steel
buildings collapsed
•Most heavily damaged
steel buildings constructed
before Japan‟s current
design code adopted (1981)
•But, even modern steel
buildings showed
unexpected damage,
including fractures at
welded beam-to-column
connections
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Good Track Record?
• „Recent‟ earthquakes (1985 Mexico City;
1994 Northridge; 1995 Hyogoken-
Nanbu) have exposed problems with
modern welded steel structures
• Care in the design, detailing, and
construction of steel structures needed
to assure satisfactory performance
• This has led to the development of
building code regulations that
specifically address seismic detailing of
steel building structures.
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Earthquakes
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• Structural Engineers Association of
California (SEAOC) Blue Book – 1988:
First comprehensive detailing provisions
for steel
• American Institute of Steel Construction
(AISC) Seismic Provisions
– 1st ed. 1990
– 2nd ed. 1992
– 3rd ed. 1997
» Supplement No. 1: February 1999
» Supplement No. 2: November 2000
– 4th ed.2002
– 5th ed.2005
– 6th ed. 2010
US Seismic Code Provisions for Steel
Northridge
& Kobe
research
findings
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Earthquakes
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Conventional Building Code Philosophy
Objective: Prevent collapse in the extreme
earthquake likely to occur at a
building site.
Objectives are not to:
- limit damage
- maintain function
- provide for easy repair
Prevent
loss of life
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Maximum Considered Earthquake
• “extreme earthquake” = Maximum
Considered Earthquake (MCE)
– In the western U.S., MCE based on the
largest earthquake that can be generated
by known faults
– In the rest of the U.S., MCE defined as an
earthquake with a 2-percent probability of
exceedance in 50 years
• recurrence interval of about 2500 years
• In MCE, can expect substantial and
costly damage to the structure
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Engelhardt’s Car Analogy
In the event of a major collision, the
design goal is to protect the occupants
of the car; not to protect the car itself.
In the event of a major earthquake, a building
is used in a sacrificial manner to absorb the
energy of the earthquake, in order to prevent
collapse and protect the occupants.
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The key to an economical design for a building
which must withstand a very strong earthquake?
Design for Ductile Behavior
HIGH
STRENGTH? Let me know if you can find
“ductile burrito” video clip!
DUCTILITY?
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H
H
Ductility = Inelastic Deformation
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H
Δyield Δfailure
Ductility Factor μ =
Δfailure
Δyield
H
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H
Strength
Required
Ductility
MAX
Helastic
3/4 *Helastic
1/2 *Helastic
1/4 *Helastic
H
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Strength
Required
Ductility
H
MAX
Helastic
3/4 *Helastic
1/2 *Helastic
1/4 *Helastic
•Trade-off between
strength and ductility
•Ductility means damage
•For a structure designed
to yield in an earthquake,
the maximum lateral
force that the structure
will see during the
earthquake is defined by
its own lateral strength
•A typical code-based
design uses ductility
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H
Ductility = Yielding
Failure =
Fracture
or
Instability
Ductility in Steel Structures: Yielding
Nonductile Failure Modes: Fracture or Instability
WILL NOT COLLAPSE
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• Choose frame elements ("fuses") that will
yield in an earthquake; e.g. beams in
moment resisting frames, braces in
concentrically braced frames, links in
eccentrically braced frames, etc.
Developing Ductile Behavior
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• Detail "fuses" to sustain large inelastic
deformations prior to the onset of fracture
or instability (i.e. , detail fuses for ductility).
M
q
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• Design frame elements to be stronger than
the fuses, i.e., design all other frame
elements to develop the plastic capacity of
the fuses.
CAPACITY DESIGN CONCEPT
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(a) (b)
Less Ductile Behavior
Ductility of Steel Frames
More Ductile Behavior
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Ductility of Steel Frames – “Backbone” Curve
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Key Elements of Seismic-Resistant Design
Lateral Forces - Strength
& Stiffness
 ASCE-7 (Minimum Design Loads
for Buildings and Other
Structures)
 National Earthquake Hazards
Reduction Program (NEHRP)
Provisions
Detailing Requirements -
Ductility
 AISC Seismic Provisions
H
H
Ductility = Inelastic
Deformation
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Design EQ Loads – Base Shear per ASCE 7-10:
V
S I
R
W
T R
W
DS
= 
S I
D1
Strength
Required
Ductility
response modification coefficient
What does it mean if R = 1.0?
R> 1.0?
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R factors for Selected Steel Systems (ASCE 7):
SMF (Special Moment Resisting Frames): R = 8
IMF (Intermediate Moment Resisting Frames): R = 4.5
OMF (Ordinary Moment Resisting Frames): R = 3.5
H
MAX
Helastic
3/4 *Helastic
1/2 *Helastic
1/4 *Helastic
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SMF (Special Moment Resisting Frames): R = 8
IMF (Intermediate Moment Resisting Frames): R = 4.5
OMF (Ordinary Moment Resisting Frames): R = 3.5
EBF (Eccentrically Braced Frames): R = 8
SCBF (Special Concentrically Braced Frames): R = 6
OCBF (Ordinary Concentrically Braced Frames): R = 3.25
BRBF (Buckling Restrained Braced Frame): R = 8
SPSW (Special Plate Shear Walls): R = 7
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Undetailed Steel Systems in
Seismic Design Categories A,
or B or C with R = 3
This availability of this option reflects the
view that a steel structure, even without
special seismic detailing, will generally exhibit
some reasonable degree of ductility.
AISC Seismic Provisions not needed;
follow main AISC specification
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R-factors
• How were
current R-factors
determined?
• R-factors for
new systems?
– ATC-63 project
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http://peer.berkeley.edu/tbi/wp-content/uploads/2010/09/Heintz_ATC-63.pdf
Some background:
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Earthquakes
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2010 AISC
Seismic
Provisions
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AISC Seismic Provisions for
Structural Steel Buildings
Symbols, Glossary, Acronyms
A. General Requirements
B. General Design Requirements
C. Analysis
D. General Member and Connection Design Requirements
E. Moment-Frame Systems
F. Braced-Frame and Shear-Wall Systems
G. Composite Moment-Frame Systems
H. Composite Braced-Frame and Shear-Wall Systems
cont’d
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I. Fabrication and Erection
J. Quality Control and Quality Assurance
K. Prequalification and Cyclic Qualification Testing Provisions
Commentary A-K
References
Structural Steel Buildings, cont’d
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A1. Scope
A2. Referenced Specifications, Codes and Standards
A3. Materials
A4. Structural Design Drawings and Specifications
B1. General Seismic Design Requirements
B2. Loads and Load Combinations
B3. Design Basis (Required Strength/Available Strength)
B4. System Type
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C. Analysis
C1. General Requirements
C2. Additional Requirements
C3. Nonlinear Analysis
D. General Member and Connection Design Requirements
D1. Member Requirements
D2. Connections
D3. Deformation Compatibility of Non-SFRS Members and
Connections
D4. H-Piles
New chapter,
more of a
“pointer” to other
sections and
documents
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Earthquakes
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2010 AISC Seismic Provisions
General Provisions Applicable to All
Systems
Highlights of Glossary
and Chapters A-D
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AISC Seismic Provisions:
Glossary - Selected Terms
Applicable Building Code (ABC)
ABC = Building code under which the structure is
designed (the local building code that
governs the design of the structure)
Where there is no local building code - use ASCE 7
We will use ASCE 7 in this course.
(Int’l Bldg Code (IBC), referenced by
Indiana Building Code, takes seismic
design requirements from ASCE 7)
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Seismic Force Resisting System (SFRS)
That part of the structural system that has been considered in
the design to provide the required resistance to the seismic
forces prescribed in ASCE/SEI 7.
Assembly of structural elements in the building that resists
seismic loads, including struts, collectors, chords,
diaphragms and trusses
www.atcouncil.org/pdfs/bp1d.pdf
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Use or Occupancy of Buildings and
Structures
Risk Category
Essential facilities
(Hospitals, fire and police stations, emergency shelters, etc)
Structures containing extremely hazardous materials
IV
Structures that could pose a substantial hazard to human
life, substantial economic impact, and/or mass disruption
of day-to-day civilian life in the event of failure
(previously defined as buildings with large assembly areas,
etc., could include facilities with hazardous materials)
III
Buildings not in Risk Categories I, III, or IV
(most buildings)
II
Buildings that represent a low risk to human life in the
event of failure
(agricultural facilities, temporary facilities, minor storage
facilities)
I
Risk Category – classification as specified by
applicable building code (ASCE 7)
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Seismic Design
Category (SDC):
ASCE 7
Classification
assigned to a
building by the
applicable building
code based upon its
risk category and the
design spectral
response
acceleration
coefficients.
Glossary
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Seismic Design Category (SDC)
SDCs:
A
B
C
D
E
F
Increasing seismic
risk
and
Increasingly
stringent seismic
design and detailing
requirements
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To Determine the Seismic Design Category (ASCE 7-10):
Determine Risk Category
Determine SS and S1
SS = spectral response acceleration for maximum considered earthquake at short periods
S1 = spectral response acceleration for maximum considered earthquake at 1-sec period
Ss and S1 are read from maps
Determine Site Class
Site Class depends on soils conditions - classified according to shear wave velocity
Determine SMS and SM1
Spectral response accelerations for maximum considered earthquake
adjusted for the Site Class;
SMS = Fa Ss SM1 = Fv S1
Fa and Fv depend on Site Class and on Ss and S1
Determine SDS and SD1
Design spectral response accelerations
SDS = 2/3 x SMS SD1 = 2/3 x SM1
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Map for S1 (ASCE 7)
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Seismic Hazard Maps
• Interactive program available from USGS website.
– Seismic design values for buildings
– Input longitude and latitude at site, or zip code
– Output SS and S1
• http://earthquake.usgs.gov/research/hazmaps/design/
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To Determine the Seismic Design Category (ASCE 7-10):
Evaluate Seismic Design Category according to
Tables 11.6-1 and 11.6-2;
The Seismic Design Category is the more severe value based on
both Tables.
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For sites with S1 ≥ 0.75g: Seismic Design Category = E for I, II, or III
Seismic Design Category = F for IV
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A1. Scope
A2. Referenced Specifications, Codes and Standards
A3. Materials
A4. Structural Design Drawings and Specifications
B1. General Seismic Design Requirements
B2. Loads and Load Combinations
B3. Design Basis (Required Strength/Available Strength)
B4. System Type
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Section A1. Scope
The Seismic Provisions shall govern the design,
fabrication and erection of structural steel members and
connections in the seismic force resisting systems
(SFRS), and splices and bases of columns in gravity
framing systems of buildings and other structures with
moment frames, braced frames and shear walls.
The Seismic Provisions are used in conjunction
with the AISC Specification for Structural Steel
Buildings
Both are in Unified LRFD-ASD format
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Use of Seismic Provisions is mandatory for
Seismic Design Category D, E or F.
Use of Seismic Provisions are mandatory for
Seismic Design Categories B or C, when using
R > 3
For Seismic Design Categories B or C: can design
using R=3 and provide no special detailing (just
design per main AISC Specification)
SDC A designed following ASCE 7 Section 1.4; AISC
Seismic Provisions do not apply.
Section A1. Scope (cont’d.)
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Section B1. General Seismic Design Requirements
Go to the Applicable Building Code for:
• Seismic Design Category
• Risk Categories
• Limits on Height and Irregularity
• Drift Limitations
• Required Strength
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Section B2. Loads and Load Combinations
Go to the Applicable Building Code
Section B3.1 Required Strength
Greater of
1) as determined by analysis, or
2) as determined by AISC Seismic Provisions
Chapter C. Analysis
Follow requirements of Applicable Building Code,
AISC Seismic Provisions, AISC Specification;
nonlinear analysis per Chapter 16 of ASCE 7
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Basic LRFD Load Combinations (ASCE-7):
1.4D
1.2D + 1.6L + 0.5(Lr or S or R)
1.2D + 1.6(Lr or S or R) + (L or 0.5W)
1.2D + 1.0W + L + 0.5(Lr or S or R)
0.9D + 1.0W
1.2D + 1.0E + L + 0.2S
0.9D + 1.0E
Load Combinations
Including E
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Definition of E for use in basic load combinations:
For Load Combination: 1.2D + 1.0E + L + 0.2S
E = ρ QE + 0.2 SDS D
For Load Combination: 0.9D + 1.0E
E = ρ QE - 0.2 SDS D
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E = ρ QE  0.2 SDS D
effect of horizontal forces effect of vertical forces
E = the effect of horizontal and vertical
earthquake-induced forces
QE = effect of horizontal earthquake-
induced forces
SDS = design spectral acceleration at short
periods
D = dead load effect
ρ = reliability factor
(depends on extent of redundancy in the
seismic lateral resisting system;
ρ varies from 1.0 to 1.3)
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Substitute E into basic load combinations:
substitute: E = ρ QE + 0.2 SDS D
substitute: E = ρ QE - 0.2 SDS D
(1.2 + 0.2 SDS) D + 1.0 ρ QE + L +0.2S
(1.2 - 0.2 SDS) D + 1.0 ρ QE
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B2. Loads and Load Combinations (cont’d.)
Where amplified seismic loads are required by
the AISC Seismic Provisions:
The horizontal portion of the earthquake load E
shall be multiplied by the overstrength factor o
prescribed by the applicable building code.
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Definition of Amplified Seismic Load (ASCE-7)
E = Ωo QE + 0.2 SDS D
Amplified Seismic Load:
E = Ωo QE - 0.2 SDS D
Amplified Seismic Load:
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Basic load combinations incorporating Amplified
Seismic Load:
substitute: E = Ωo QE + 0.2 SDS D
substitute: E = Ωo QE - 0.2 SDS D
(1.2 + 0.2 SDS) D + Ωo QE + L +0.2S
(0.9 - 0.2 SDS) D + Ωo QE
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Seismic Overstrength Factor: Ωo
System Ωo
Moment Frames (SMF, IMF, OMF)
Concentrically Braced Frames (SCBF, OCBF)
Eccentrically Braced Frames (EBF)
Special Plate Shear Walls (SPSW)
Buckling Restrained Braced Frames (BRBF)
3
2
2
2
2.5
Per ASCE-7:
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Amplified Seismic Load
Lateral
Seismic
Force
Frame Lateral Deflection
Qe
Ωo Qe
Amplified Seismic Load, ΩoQe, is intended to provide an
estimate of a frame's plastic lateral strength
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Amplified Seismic Load, cont’d
Lateral
Seismic
Force
Frame Lateral Deflection
Qe
Ωo Qe
•Reasons for
overstrength
– Use of resistance factors
– Actual yield stress
– Members sized to satisfy
drift limits
– Members sized to
simplify design and
construction
– Increase in strength in
going from 1st plastic hinge
to plastic mechanism
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Section A3.1 Material Specifications
Limits and ASTM Specifications
Section A3.2 Expected Material Strength
For determining required strength as applicable
Section A3.3 Heavy Sections
Toughness requirements
Section A3.4 Consumables for Welding
SFRS, Demand Critical welds (discuss more later)
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A3.1 Material Specifications
For members in which inelastic behavior is
expected:
Specified minimum Fy ≤ 50 ksi
Exceptions:
• Columns for which only expected yielding
is at the base
• Members in OMFs, OCBFs , C-OMFs, C-
OBFs, C-OSWs (permitted to use up to Fy
= 55 ksi)
Grade 65 can be advantageous
To accommodate
materials commonly used
in metal building systems
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88
A3.1 Material Specifications
For members in which inelastic behavior is
expected:
Specified minimum Fy ≤ 50 ksi
WHY?
Majority of experiments conducted
on seismic frame elements has
been for steels with specified yield
stress of 50 ksi and less.
Higher strength steels tend to be
more brittle.
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89
A3.2 Expected Material Strength
Expected Yield Strength = Ry Fy
Expected Tensile Strength = Rt Fu
Fy = minimum specified yield strength
Fu = minimum specified tensile strength
Ry and Rt are based on statistical analysis of
mill data.
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90
Ry
Rt
Added to Seismic Provisions after 1994
Northridge Earthquake
Added to Seismic Provisions more recently
for checks of fracture limit states in same
member for which expected yield stress is
used (motivated by Braced Frame design)
connections
1.1RyFyZ
connections
RyFyAg
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91
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92
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93
Example: A36 angles used for brace in an SCBF
Fy = 36 ksi
Fu = 58 ksi
Ry Fy = 1.5  36 ksi = 54 ksi
Rt Fu = 1.2  58 ksi = 70 ksi
Example: A992 wide flange used for beam in an SMF
Fy = 50 ksi
Fu = 65 ksi
Ry Fy = 1.1  50 ksi = 55 ksi
Rt Fu = 1.1  65 ksi = 72 ksi
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94
Where specified in the Seismic Provisions, the
required strength of a member or connection shall
be based on the Expected Yield Strength, Ry Fy of
an adjoining member.
The Expected Tensile Strength, Rt Fu and the
Expected Yield Strength, Ry Fy may be used to
compute the nominal strength for rupture and
yielding limit states within the same member.
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95
Example: SCBF Brace and Brace Connection
To size brace member:
Required Strength defined by
code specified forces (using
ASCE-7 load combinations)
Design Strength of member
computed using minimum
specified Fy
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96
Example: SCBF Brace and Brace Connection (cont)
Required Axial Tension Strength of
brace connection is the expected
yield strength of bracing member =
Ry Fy Ag
Ry Fy Ag
Note: no 1.1 multiplier for strain
hardening (used for moment
connections); braces exhibit little
strain hardening
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97
Ry Fy Ag
Gusset Plate:
Compute design strength
using minimum specified Fy
and Fu of gusset plate material
Design strength
should exceed
Required Axial Tension
Strength of brace
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98
Ry Fy Ag
Bolts:
Compute design shear
strength using minimum
specified Fu of bolt
Design strength
should exceed
Required Axial Tension
Strength of brace
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99
Ry Fy Ag
Net Section Fracture and
Block Shear Fracture of
Bracing Member:
Compute design strength
using expected yield
strength, RyFy and expected
tensile strength, Rt Fu of the
brace material.
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100
Ry Fy Ag
For example:
The required design
strength for limit states
of net section fracture
and block shear is
RyFyAg.
Block shear fracture:
[Ant RtFu + 0.6 AnvRtFu] ≤ [Ant RtFu + 0.6AgvRyFy]
Net section fracture:
 AeRtFu
Whenever the required strength is based on the expected yield strength of
an element, then the design strength of that same element can be computed
using expected yield and tensile strength.
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101
Section D1.3 Member Requirements: Protected Zones
Section D2.1 Connections: General
Section D2.2 Bolted Joints
Section D2.4 Continuity Plates and Stiffeners
Section D2.5 Column Splices
Section D2.6 Column Bases
Section D2.3 Welded Joints
Start here
and then
discuss
D1.3
Revisit Section A3.4 here
Discuss
with
“Members”
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102
D2.1 Connections: General
Connections, joints and fasteners that are
part of the seismic force resisting system
(SFRS) shall comply with the AISC
Specification Chapter J, and with the
additional requirements in this section.
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103
D2.2 Bolted Joints
Requirements for bolted joints:
• All bolts must be high strength (A325 or A490)
• Bolted joints may be designed as bearing type connections,
but must be constructed as slip critical
- bolts must be pretensioned
- faying surfaces must satisfy Class A surface reqs.
• Holes: standard size or short-slots perpendicular to load
(exception: oversize holes are permitted for diagonal brace
connections, but the connection must be designed as slip-
critical and the oversize hole is permitted in one ply only)
• Nominal bearing strength at bolt holes shall not be taken as
greater than 2.4 d t Fu
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104
D2.2 Bolted Joints
Bolts and welds shall not be
designed to share force in a
joint, or the same force
component in a connection.
Not Permitted
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105
Fig. C-D2.1. Desirable details that avoid shared forces between welds and bolts.
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106
Fig. C-D2.1. Desirable details that avoid shared forces between welds and bolts.
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107
D2.3 Welded Joints
Welded joints shall be designed in accordance with
Chapter J of the Specification.
All welds used in members and connections in the
SFRS shall be made with filler metals meeting the
requirements specified in clause 6.3 of Structural
Welding Code—Seismic Supplement (AWS
D1.8/D1.8M).
A3.4a Seismic Force Resisting System Welds
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108
A3.4a Seismic Force Resisting System Welds
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109
A3.4b Demand Critical Welds
Demand Critical – subjected to very high demands;
specifically identified in the Provisions in section
applicable to designated SFRS
Must ALSO satisfy:
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110
D1.3 Protected Zone
Discontinuities specified in Section I2.1 resulting from
fabrication and erection procedures and from other
attachments are prohibited in the area of a member or a
connection element designated as a protected zone by
these Provisions or ANSI/AISC 358.
Exception: Welded steel headed stud anchors and other
connections are permitted in protected zones when
designated in ANSI/AISC 358, or as otherwise determined
with a connection prequalification in accordance with
Section K1, or as determined in a program of qualification
testing in accordance with Sections K2 and K3.
Attachments in the highly strained protected zones may
serve as fracture initiation sites
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111
D1.3 Protected Zone
Some examples of prohibited attachments/
discontinuities:
• No welded shear studs are permitted.
• No decking attachments that penetrate the beam flange
are permitted (no powder actuated fasteners); but, decking
arc spot welds are permitted.
• No welded, bolted, screwed, or shot-in attachments for
edge angles, exterior facades, partitions, duct work, piping,
etc are permitted.
• Discontinuities from fabrication or erection operations
(such as tack welds, erection aids, etc) shall be repaired.
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112
Examples of Protected Zones: SMF
Protected Zones
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113
Examples of Protected Zones: SCBF
Protected Zones
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114
Examples of Protected Zones: EBF
Protected Zones
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115
Section D1.1 Member Requirements: Classification of
Sections for Ductility
Section D1.1a Section Requirements for Ductile
Members
Section D1.1b Width-to-Thickness Limitations of Steel
and Composite Sections
Section D2.5 Column Splices
Section D2.6 Column Bases
Section D1.4 Columns
Go back to:
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116
Section D1.1 Member Requirements: Classification of
Sections for Ductility
Local buckling of members can significantly affect both
strength and ductility of the member.
When required for the systems defined in Chapters E, F,
G, H and Section D4, members designated as
moderately ductile members or highly ductile members
shall comply with this section.
Plastic rotation
0.02 rad or less
Plastic rotation
0.04 rad or more
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117
Section D1.1a Section Requirements for Ductile Members
Structural steel sections for both moderately ductile
members and highly ductile members shall have
flanges continuously connected to the web or webs.
Section D1.1b Width-to-Thickness Limitations of Steel and
Composite Sections
For members designated as moderately ductile
members, the width-to-thickness ratios of compression
elements shall not exceed the limiting width-to-thickness
ratios, λmd, from Table D1.1.
For members designated as highly ductile members, the
width-to-thickness ratios of compression elements shall
not exceed the limiting width-to-thickness ratios, λhd,
from Table D1.1.
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118
Local buckling of a moment frame beam.....
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119
Local buckling of an EBF link.....
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120
Local buckling of an HSS column....
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121
Local buckling of an HSS brace.....
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122
M
q
Mp
Increasing b / t
Effect of Local Buckling on Flexural Strength and Ductility
M
q
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123
Mr
Moment
Capacity
md r
Width-Thickness Ratio
Mp
Plastic Buckling
Inelastic Buckling
Elastic Buckling
hd
Ductility
Effect of Local Buckling on Flexural Strength and Ductility
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124
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125
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126
D1.4a Columns: Required Strength
The required strength of columns in the SFRS shall be
determined from the following:
(1) The load effect resulting from the analysis
requirements for the applicable system
(2) The compressive axial strength and tensile strength as
determined using the load combinations stipulated in
the applicable building code including the amplified
seismic load. It is permitted to neglect applied moments
in this determination unless the moment results from a
load applied to the column between points of lateral
support.
(1.2 + 0.2 SDS) D + Ωo QE + L +0.2S
(1.2 - 0.2 SDS) D + Ωo QE
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127
D1.4a Columns: Required Strength
The required axial compressive strength and tensile
strength need not exceed either of the following:
(a) The maximum load transferred to the column by the
system, including the effects of material overstrength and
strain hardening in those members where
yielding is expected.
(b) The forces corresponding to the resistance of the
foundation to overturning uplift.
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D2.5 Column Splices
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D2.5 Column Splices
Column splices in any SFRS frame must satisfy
requirements of Section D1.4a (Required
Strength for Columns)
Additional requirements for columns splices are
specified for:
- Moment Frames (Chapter E)
- Braced Frames and Shear Walls (Chapter F)
- Composite Braced-Frame and Shear-Wall Systems
(Chapter H)
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D2.5 Column Splices
The required strength
determined using the load
combinations stipulated in
the applicable building
code including the
amplified seismic load.
The required strength
need not exceed the
maximum loads that can
be transferred to the splice
by the system.
Pu - splice
Mu - splice
Vu - splice
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D2.5 Column Splices
Welded column splices subjected to net
tension when subjected to amplified
seismic loads, shall satisfy both of the
following requirements:
1. If partial joint penetration (PJP) groove
welded joints are used, the design strength of
the PJP welds shall be at least 200-percent of
the required strength.
And....
2. The design strength of each flange splice
shall be at least 0.5 Ry Fy Af for the smaller
flange
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D2.5 Column Splices
PJP Groove Weld
Stress concentration:
Fracture initiation
point.
Design PJP groove
weld for 200 % of
required strength
( PJP Groove welds not permitted in column splices
for Special and Intermediate Moment Frames)
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D2.5 Column Splices
For all building columns including those
not designated as part of the SFRS, the
required shear strength of column splices
with respect to both orthogonal axes of the
column shall be Mpc/H (LRFD), where Mpc is
the lesser nominal plastic flexural strength
of the column sections for the direction
in question, and H is the height of the story.
The required shear strength of splices of
columns in the SFRS shall be the greater of
the above requirement or the required
shear strength determined per Section
D2.5b(a) and (b).
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D2.5 Column Splices
4 ft. min
Splices made with fillet
welds or PJP welds shall
be located at least 4-ft.
from beam-to-column
connections
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Moment Resisting Frames
• Definition and Basic Behavior of Moment
Resisting Frames
• Beam-to-Column Connections: Before and After
Northridge
• Panel-Zone Behavior
• AISC Seismic Provisions for Moment Resisting
Frames: Special, Intermediate and Ordinary
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MOMENT RESISTING FRAME (MRF)
Advantages
• Architectural Versatility
• High Ductility and Safety
Disadvantages
• Low Elastic Stiffness
Beams and columns with moment resisting
connections; resist lateral forces by flexure and
shear in beams and columns
Develop ductility by:
- flexural yielding of beams
- shear yielding of column panel zones
- flexural yielding of columns
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Achieving Ductile Behavior:
• Choose frame elements ("fuses") that will
yield in an earthquake, i.e, choose plastic
hinge locations.
• Detail plastic hinge regions to sustain
large inelastic rotations prior to the onset
of fracture or instability.
• Design all other frame elements to be
stronger than the plastic hinge regions.
Understand and Control Inelastic Behavior:
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Behavior of an MRF Under Lateral Load:
Internal Forces and Possible Plastic Hinge Locations
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M V
Note high shear (V) in Panel Zones
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Possible Plastic Hinge Locations
Beam
(Flexural Yielding)
Panel Zone
(Shear Yielding)
Column
(Flexural & Axial
Yielding)
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Plastic
Hinges
In Beams
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Plastic Hinges
In Column
Panel
Zones
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Plastic Hinges
In Columns:
Potential for
Soft Story
Collapse
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Critical Detailing Area for Moment Resisting Frames:
Beam-to-Column Connections
Design
Requirement:
Frame must develop
large ductility
without failure of
beam-to-column
connection.
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Moment Connection Design Practice Prior to
1994 Northridge Earthquake:
Welded flange-bolted
web moment connection
widely used from early
1970’s to 1994
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Moment Connection Design Practice Prior to
1994 Northridge Earthquake:
Considered less desirable because of
slip of bolts (pinched hysteresis loops)
and net section rupture
All-bolted
connection?
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Pre-Northridge
Welded Flange – Bolted Web Moment Connection
Backup Bar
Beam Flange
Column Flange
Stiffener
Weld Access Hole
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Welded Flange –
Bolted Web
Moment
Connection
Weld tabs in
place
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Stages of
construction of
welded flange –
bolted web
moment
connection.
Beam web bolted
to shear tab.
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Typical: 3/8” root
30-degree bevel on beam
flange
Bottom flange
back-up bar tack
welded into place.
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Weld tabs tack
welded into place,
extending groove
geometry beyond
flange edges.
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First weld pass
has been placed
with flux-cored arc
welding (FCAW)
process.
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Bottom groove
continues to be
filled.
Note interruption at
middle portion of
flange.
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Completed bottom
flange groove
weld.
Pre-Northridge
practice: back-up
bar and weld tabs
left in place.
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Beam top flange
with back-up bar
and weld tabs in
place.
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Completed top
flange groove
weld.
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Ultrasonic Testing
(UT) on a moment
connection with a
cover plate.
UT used to detect
defects.
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Experimental Data on “Pre-Northridge”
Moment Connection
Typical Experimental
Setup
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Initial Tests on Large Scale Specimens:
• Tests conducted at
UC Berkeley ~1970
• Tests on W18x50 and
W24x76 beams
• Tests compared all-
welded connections
with welded flange-
bolted web
connections
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Note on initial test specimens
Loss of redundancy, deeper beams, thicker
column flanges!
Relatively modest beam and column sizes
At the time, nearly all beam-to-column
connections in buildings designed to transfer
moment
Over the years, cost premium for full moment
connections led engineers to limit number of
bays of framing designed as ductile moment-
resisting frames
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All-Welded Detail
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Welded Flange – Bolted Web Detail
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Observations from Initial UC Berkeley Tests:
• Large ductility developed by all-welded
connections.
• Welded flange-bolted web connections
developed less ductility, but were
viewed as still acceptable.
At that time (early 1970s), little
information available on level of
ductility needed to survive a
strong earthquake.
Welded flange – bolted web connection beam the “de facto
standard”, used in a large number of moment frames
Less costly to
fabricate!
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Subsequent Test Programs:
• Welded flange-bolted web connections
showed highly variable performance.
• Typical failure modes: fracture at or
near beam flange groove welds.
• A large number of laboratory tested
connections did not develop adequate
ductility in the beam prior to connection
failure.
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Bottom flange groove weld fracture
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Top flange fracture, initiated at left edge
at weld-runoff region.
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Summary of Testing Prior to
• Welded flange – bolted web connection
showed highly variable performance
– Identical specimens (different welder),
welds inspected – vast difference in
demonstrated ductility or lack thereof
(Engelhardt and Hussain, 1993)
• Many connections failed in laboratory with
little or no ductility
Reasons not well understood.
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Widespread
failure of
welded flange -
bolted web
moment
connections
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• January 17, 1994
• Magnitude = 6.8
• Epicenter at Northridge - San Fernando
Valley
(Los Angeles area)
• Fatalities: 58
• Estimated Damage Cost: $20 Billion
(structural and non-structural)
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Northridge - Ground Accelerations
• Sylmar: 0.91g H 0.60g V
• Sherman Oaks: 0.46g H 0.18g V
• Granada Hills: 0.62g H 0.40g V
• Santa Monica: 0.93g H 0.25g V
• North Hollywood: 0.33g H 0.15g V
@alirezasalehin
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Collapse of first
story of a wood-
framed
apartment
building.
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Nonductile reinforced
concrete frame building
(collapse of entire story)
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Modern precast
parking garage
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Damage to Steel Buildings in the
• Initially not identified (not readily visible); found
accidentally later during repairs to nonstructural
elements, observations of elevator problems, etc.
• Large number (more than 100 of approx. 500 in
region) of modern steel buildings sustained severe
damage at beam-to-column connections.
• Primary Damage: Fracture in and around beam flange
groove welds
• Damage was largely unexpected by engineering
profession
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Backup Bar
Beam Flange
Column Flange
Stiffener
Weld Access Hole
Pre-Northridge
Welded Flange – Bolted Web Moment Connection
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Weld Tab
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“Divot” failure
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Divot type fracture
(laboratory test
specimen)
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Damage Observations
• A large number of steel moment frame
buildings suffered connection damage
• No steel moment frame buildings collapsed
• Typical Damage:
– fracture of groove weld
– “divot” fracture within column flange
– fracture across column flange and web
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Observations from Studies of
Fractured Connections
• Many connections failed by brittle fracture with little
or no ductility
• Brittle fractures typically initiated in beam flange
groove welds
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Response to Northridge Moment Connection Damage
(S) Structural Engineers Association of California
(A) Applied Technology Council
(C) California Universities for Research in Earthquake Engineering
• Nearly immediate elimination of welded
flange - bolted web connection from US
building codes and design practice
• Intensive research and testing efforts to
understand causes of damage and to develop
improved connections
– AISC, NIST, NSF, etc.
– SAC Program (FEMA)
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Causes of Moment Connection Damage in
Northridge
• Welding
• Connection Design
• Materials
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Causes of Northridge Moment
Connection Damage:
Welding Factors
• Low Fracture Toughness of Weld
Metal
• Poor Quality
• Effect of Backing Bars and Weld Tabs
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Weld Metal Toughness
Most common Pre-
Northridge welding
electrode (E70T-4)
had very low
fracture
toughness.
Typical Charpy V-
Notch: < 5 ft.-lbs at
700F
(7 J at 210C)
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Welding Quality
• Many failed connections showed evidence of
poor weld quality
• Many fractures initiated at root defects in
bottom flange weld, in vicinity of weld access
hole
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Lack of penetration defect
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Weld Backing Bars and Weld Tabs
• Backing Bars:
– Can create notch effect
– Increases difficulty of inspection
• Weld Tabs:
– Weld runoff regions at weld tabs contain numerous
discontinuities that can potentially initiate fracture
@alirezasalehin
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Note: shows
evidence of
lamellar
tearing
@alirezasalehin
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Design Factors:
Stress/Strain Too High at Beam Flange Groove Weld
• Inadequate Participation of Beam Web Connection in
Transferring Moment and Shear
• Effect of Weld Access Hole
• Effect of Column Flange Bending
• Other Factors
Causes of Northridge Moment Connection
Damage:
Including presence of composite floor slab
Panel Zone (more later)
@alirezasalehin
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Mp
Increase in Flange Stress Due to
Inadequate Moment Transfer Through Web Connection
Flange
Stress
Fy
Fu
@alirezasalehin
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Vflange
Increase in Flange Stress Due to Shear in Flange
@alirezasalehin
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Stress
Concentrations:
• Weld access
hole
• Shear in flange
• Inadequate
flexural
participation of
web connection
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Base Metal
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Tri-Axial Stress Condition
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Causes of Moment Connection Damage in
Northridge:
Material Factors (Structural Steel)
• Actual yield stress of A36 beams often
significantly higher than minimum
specified
@alirezasalehin
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FEMA 267, August 1995
Interim Guidelines: Evaluation,
Repair, Modification and
Design of Welded Steel
Moment Frame Structures
Advisory No. 1 – 1997
Advisory No.2 - 1999
@alirezasalehin
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Strategies for Improved Performance of
Moment Connections
• Welding
• Materials
• Connection Design and Detailing
@alirezasalehin
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Strategies for Improved Performance of Moment
Connections:
WELDING
• Required minimum toughness for weld metal:
– Required CVN for all welds in SFRS:
20 ft.-lbs at 00 F
– Required CVN for Demand Critical welds:
40 ft.-lbs at 700 F
ANSI/AISC 341-10 Section A3.4
@alirezasalehin
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WELDING
• Improved practices for backing bars and weld tabs
Typical improved practice:
– Remove bottom flange backing bar
– Seal weld top flange backing bar
– Remove weld tabs at top and bottom flange welds
• Greater emphasis on quality and quality control (AISC
Seismic Provisions – Chapter J)
Connections:
@alirezasalehin
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Typical Pre-Northridge
Bottom Flange Weld
@alirezasalehin
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Improved POST-Northridge Bottom Flange Weld
Weld tabs and
runoff regions
removed; ground
smooth
Back-up bar removed; root visually
inspected, defects removed; small
reinforcing fillet weld placed at
bottom of groove weld
@alirezasalehin
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Typical Pre-Northridge
Top Flange Weld
@alirezasalehin
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Improved POST-
Northridge Top
Flange Weld
Weld tabs and
runoff regions
removed;
ground
smooth
Back-up bar left in place; small
fillet weld placed between bar
and column face
@alirezasalehin
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Connections:
Materials (Structural Steel)
• Introduction of “expected yield stress” into design
codes
Fy = minimum specified yield strength
Ry = 1.5 for ASTM A36 (hot-rolled shapes and bars)
= 1.1 for A572 Gr. 50 and A992
Expected Yield Stress = Ry Fy
ANSI/AISC 341-10 Table A3.1
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Connections:
Materials (Structural Steel)
• Introduction of ASTM A992 steel for wide flange
shapes
ASTM A992
Minimum Fy = 50 ksi
Maximum Fy = 65 ksi
Minimum Fu = 65 ksi
Maximum Fy / Fu = 0.85
@alirezasalehin
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Improved Weld Access
Hole
From 2005 AISC Seismic
Provisions, with
dimensions and finish
requirements
ANSI/AISC 358-10 Section 8.5
Weld access hole
geometry (and quality!)
shall conform to
requirements of AWS D1.8.
@alirezasalehin
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Improved Weld Access
Hole
Notes:
1. Bevel as required for selected
groove weld.
2. Larger of tbf or ½ in. (13 mm) (plus ½
tbf, or minus ¼ tbf)
3. ¾ tbf to tbf, ¾ in. (19 mm) minimum (
¼ in.) ( 6 mm)
4. 3/8 in. (10 mm) minimum radius (plus
not limited, minus 0)
5. 3 tbf ( ½ in.) (13 mm)
Tolerances shall not accumulate to the
extent that the angle of the access
hole cut to the flange surface
exceeds 25.
(from 2005 Seismic Design Manual)
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Pre-Northridge
Improved
@alirezasalehin
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Connections:
Connection Design
• Development of Improved Connection Designs
and Design Procedures
– Reinforced Connections
– Proprietary Connections
– Reduced Beam Section (Dogbone)
Connections
– Other SAC Investigated Connections
@alirezasalehin
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Reinforced Connection
Cover-Plated
Connection
Cover plates fillet welded
to beam flanges, then
combined beam flange
and cover plate groove
welded to face of column
@alirezasalehin
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Cover-Plated Connection
Improved
performance in
general, but
costly to
construct
@alirezasalehin
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Cover-Plated Connection • About 2/3 of specimens
developed total plastic
rotation of 0.03 rad
without brittle fracture
(strong panel zone)
• Others, panel zone
yielding dominated
response
• 2 specimens with bolted
webs failed in brittle
manner (< 0.02 rad
plastic rotation)
• Brittle fracture in
specimen for which
welding procedure not
enforced
• Failure in specimen with
LONG cover plate
Not sufficiently
reliable?
Susceptible to same
problems of weld
quality and through-
thickness behavior of
column flange?
@alirezasalehin
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Flange Rib
Connection
Like cover-plated connection,
connection is stronger than
beam, plastic hinge formation
forced away from face of
column
@alirezasalehin
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Flange Rib
Connection
@alirezasalehin
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Lee, C.H. et al. (2005) “Experimental Study of Cyclic Seismic Behavior of
Steel Moment Connections Reinforced with Ribs,” Journal of Structural
Engineering, Vol. 131, No. 1, January 1, 2005
@alirezasalehin
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Haunched
Connection
May be at bottom flange or
both top and bottom flanges
Initially tested as four pre-
Northridge connections
repaired with bottom
triangular T-shaped haunches
@alirezasalehin
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Haunched Connections
• Generally good
performance in
laboratory
• Costly to construct
• Not included as
„prequalified‟ for new
buildings in FEMA350
Recommended Seismic
Design Criteria for New
Steel Moment-Frame
Buildings (2000)
– More on FEMA 350 later
@alirezasalehin
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Reduced Beam
Section (RBS)
Also called “Dogbone”
connection; less costly, simpler
than reinforced connections
Forces hinge formation to
occur within reduced section
@alirezasalehin
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RBS has become one of the most common moment
connection details used in current practice.
(More details later …)
@alirezasalehin
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Proprietary Connection
“Side Plate”
Connection
Beam flanges NOT directly
welded to column flanges;
forces transferred through side
plates.
http://www.sideplate.com/
@alirezasalehin
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Proprietary Connection
Slotted Web
Connection
Slots serve to reduce stress
concentrations in beam flanges
and groove welds
Seismic Structural Design
Associates (SSDA), Inc.
http://www.slottedweb.com/
@alirezasalehin
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Connections Investigated Through
SAC-FEMA Research Program
@alirezasalehin
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Reduced Beam
Section (RBS)
@alirezasalehin
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Welded
Unreinforced
Flange - Bolted
Web (WUF-B)
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Welded
Unreinforced
Flange - Welded
Web (WUF-W)
@alirezasalehin
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Free Flange
Connection
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Welded Flange
Plate Connection
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Bolted Unstiffened
End Plate
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Bolted Stiffened
End Plate
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Bolted Flange
Plate
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Double Split Tee
Typically treated
as a partially-
restrained
connection
(effects of
connection
flexibility must
be included in
overall frame
analysis).
@alirezasalehin
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Results of SAC-FEMA Research Program
Recommended Seismic Design Criteria
for Steel Moment Frames
• FEMA 350
Recommended Seismic Design Criteria for New Steel Moment-
Frame Buildings
• FEMA 351
Recommended Seismic Evaluation and Upgrade Criteria for
Existing Welded Steel Moment-Frame Buildings
• FEMA 352
Recommended Postearthquake Evaluation and Repair Criteria
for Welded Steel Moment-Frame Buildings
• FEMA 353
Recommended Specifications and Quality Assurance
Guidelines for Steel Moment-Frame Construction for Seismic
Applications
@alirezasalehin
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FEMA 350
•Nine moment
connection details just
shown – “prequalified
connections”
•Recommended design
procedures, limits of
usage (e.g., OMF only,
W36 beams and
shallower, flange
thickness limits, web
connection, etc.)
•Not a standard; but
still a valuable
reference
@alirezasalehin
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•New standard
•Replaced FEMA 350
•Prequalified – “rigorous
program of testing, analytical
evaluation and review by … the
connection prequalification
review panel (CPRP).”
–Reduced Beam Section
(RBS)
–Bolted Stiffened and
Unstiffened Extended End
Plate
ANSI/AISC 358-05
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Connection Prequalification
• AISC Connection Prequalification
Review Panel
• ANSI/AISC 341-10 Chapter K
– “Connections shall be prequalified based on
test data satisfying Section K1.3, supported
by analytical studies and design models.
The combined body of evidence… must be
sufficient to assure that the connection can
supply the required story drift angle for
SMF and IMF systems …. on a consistent
and reliable basis within the specified limits
of prequalification….”
@alirezasalehin
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ANSI/AISC 358-10
Proprietary Connections
@alirezasalehin
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Reduced Beam Section (RBS)
@alirezasalehin
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Bolted Unstiffened Extended End Plate (BUEEP)
Bolted Stiffened Extended End Plate (BSEEP)
@alirezasalehin
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Bolted Flange Plate (BFP)
@alirezasalehin
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Welded Unreinforced Flange – Welded
Web (WUF-W)
@alirezasalehin
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Kaiser Bolted Bracket (KBB)
@alirezasalehin
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“Steel Connections: Proprietary or Public Domain?” by P. Cordova & R.
Hamburger, Modern Steel Construction, October 2011
@alirezasalehin
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Experimental Evaluation of Kaiser
Bolted Bracket Steel Moment-
Resisting Connections
Scott M. Adan and William Gibb
AISC Engineering Journal 2009
http://www.steelcastconnections.com/
@alirezasalehin
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ConXtech ConXL moment connection
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http://dcm-designs.com/steel-prefabricated-moment-frame/
“This innovative
connection system
enables beams to be
simply lowered and locked
onto square columns in
the field, resulting in a
dimensionally accurate
structural chassis. The
system is often referred to
as a full-scale erector set.”
http://www.conxtech.com
/conx-system/
VIDEO CLIP
@alirezasalehin
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Connections in process of prequalification
• Double Tee
• Simpson Strong Frame
• SENSE TSC
• Side Plate
• SOM Pin Fuse Joint
@alirezasalehin
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Double Tee
http://www.aisc.org/uploadedcontent/2012NASCCSessions/N11/
@alirezasalehin
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Simpson Strong Frame (Yield Link)
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SENSE TSC
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SidePlate
@alirezasalehin
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SOM Pin Fuse Joint
“Steel Connections: Proprietary or Public Domain?” by P. Cordova & R.
Hamburger, Modern Steel Construction, October 2011
@alirezasalehin
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SOM Pin Fuse Joint
@alirezasalehin
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http://www.som.com/content.cfm/pin_fuse_joint
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Column Panel Zone
Column Panel Zone:
- subject to high shear
- shear yielding and large
shear deformations possible
(forms “shear hinge”)
- provides alternate yielding
mechanism in a steel moment
frame
@alirezasalehin
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Plastic Shear Hinges
In Column Panel Zones
Important
Questions:
Does panel zone yielding result in ductile behavior?
Is this an acceptable approach for moment frame design?
@alirezasalehin
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A number of
experiments have
been conducted
with specimens
with weak panel
zones
@alirezasalehin
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Another
experiment
(cover-plated
connection) where
the panel zone is
the primary
yielding element.
@alirezasalehin
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Very weak panel
zone; localized
“kinks” cause
strain
concentrations,
ultimately leading
to fracture in
vicinity of beam
flange groove
welds.
@alirezasalehin
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Same specimen
as previous slide.
Connection failed
at moment well
below Mp
@alirezasalehin
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"kink" at corners
of panel zone
@alirezasalehin
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-400
-300
-200
-100
0
100
200
300
400
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Story Drift Angle (rad)
Column
Tip
Load
(kips)
Composite RBS Specimen with
Weak Panel Zone
@alirezasalehin
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-1200
-800
-400
0
400
800
1200
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Panel Zone g (rad)
Panel
Zone
Shear
Force
(kips)
Composite RBS Specimen with
Weak Panel Zone
g
@alirezasalehin
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Observations on Panel Zone Behavior
• Very high ductility is possible.
• Localized deformations (“kinking”) at corners
of panel zone may increase likelihood of
fracture in vicinity of beam flange groove
welds.
• Current AISC Seismic Provisions permits
limited yielding in panel zone (Specification
J10.6 for available strength)
• Further research needed to better define
acceptable level of panel zone yielding
@alirezasalehin
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2010 AISC Seismic Provisions
Section E3
Special Moment Frames (SMF)
Section E2
Intermediate Moment Frames (IMF)
Section E1
Ordinary Moment Frames (OMF)
@alirezasalehin
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Section E3
1. Scope
2. Basis of Design
3. Analysis
4. System Requirements
4a. Moment Ratio
4b. Stability Bracing of Beams
4c. Stability Bracing at Beam-to-Column Connections
5. Members
5a. Basic Requirements
5b. Beam Flanges
5c. Protected Zones
No additional analysis requirements
@alirezasalehin
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Section E3
6. Connections
6a. Demand Critical Welds
6b. Beam-to-Column Connections
6c. Conformance Demonstration
6d. Required Shear Strength
6e. Panel Zone
6f. Continuity Plates
6g. Column Splices
@alirezasalehin
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AISC Seismic Provisions - SMF
E3.1, E3.2 Scope & Basis of Design
Special moment frames (SMF) are expected to
withstand significant inelastic deformations
through flexural yielding of the SMF beams and
limited yielding of column panel zones.
Flexural yielding of columns at base
permitted.
@alirezasalehin
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E3.1, E3.2 Scope & Basis of Design
Design of connections of
beams to columns, including
panel zones and continuity
plates, shall be based on
connection tests that provide
the performance required by
Section E3.6b, and
demonstrate this conformance
as required by Section E3.6c.
@alirezasalehin
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E3.4a Moment Ratio
SectionE3.4a requires strong column - weak
girder design for SMF (with a few exceptions)
Purpose of strong column -
weak girder requirement:
Prevent Soft Story Collapse
The following relationship shall be satisfied at
beam-to-column connections:
0
.
1
M
M
*
pb
*
pc
Eqn. (E3-1)
@alirezasalehin
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E3.4a Moment Ratio - Exceptions
Columns with Pu < 0.3 FyAg and
i) In one-story building or top story of
multistory building
ii) Sum of available shear strengths of all
exempted columns in the story is less
than 20% of sum of available shear
strengths of all columns in that story
acting in same direction; or less than
33% if considering one column line
OR columns in any story that has ratio of available shear
strength to required shear strength that is 50% greater than
story above
@alirezasalehin
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E3.4a Moment Ratio
0
.
1
M
M
*
pb
*
pc
*
pc
M the sum of the moments in the column above and below the joint at
the intersection of the beam and column centerlines.
∑M*
pc is determined by summing the projections of the nominal
flexural strengths of the columns above and below the joint to
the beam centerline with a reduction for the axial force in the
column.
It is permitted to take ∑M*
pc = ∑Zc ( Fyc - Puc/Ag)
*
pb
M the sum of the moments in the beams at the intersection of the beam
and column centerlines.
∑M*
pb is determined by summing the projections of the expected
flexural strengths of the beams at the plastic hinge locations to
the column centerline.
@alirezasalehin
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C Column
L
C Beam
L
M*pc-top
M*pc-bottom
M*pb-left
M*pb-right
0
.
1
M
M
*
pb
*
pc
Note:
M*pc is based on minimum specified yield
stress of column
M*pb is based on expected yield stress of beam
and includes allowance for strain hardening
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Mpr-right
Mpr-left
Vbeam-right
Vbeam-left
Left Beam Right Beam
Plastic Hinge Location
sh+dcol/2
Mpr = expected moment at plastic hinge = 1.1 Ry Mp or as specified in ANSI/AISC 358
Vbeam = beam shear (see Section E3.6d - beam required shear strength)
sh = distance from face of column to beam plastic hinge location (specified in
ANSI/AISC 358)
M*pb-left M*pb-right
sh+dcol/2
M*pb = Mpr + Vbeam (sh + dcol /2 )
Computing M*pb
@alirezasalehin
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Top Column
Bottom Column
Mpc = nominal plastic moment capacity of column, reduced for presence of axial
force; can take Mpc = Zc (Fyc - Puc / Ag) [or use more exact moment-axial force
interaction equations for a fully plastic cross-section]
Vcol = column shear - compute from statics, based on assumed location of column
inflection points (usually midheight of column)
M*pc-bottom
M*pc = Mpc + Vcol (dbeam /2 )
Computing M*pc
Mpc-bottom
Mpc-top
M*pc-top
dbeam
Vcol-top
Vcol-bottom
@alirezasalehin
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E3.4b Stability Bracing of Beams
Beams shall be braced to satisfy the
requirements for highly ductile members in
Section D1.2b.
The required strength of stability bracing
provided adjacent to plastic hinges shall be as
required by Section D1.2c.
@alirezasalehin
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Lateral Torsional Buckling
Lateral torsional
buckling controlled by:
y
b
r
L
Lb = distance between beam lateral braces
ry = weak axis radius of gyration
Lb Lb
Beam lateral braces (top & bottom flanges)
@alirezasalehin
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M
Mp
Increasing Lb / ry
Effect of Lateral Torsional Buckling on Flexural Strength and Ductility:
M
@alirezasalehin
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ksi
50
F
for
r
50
r
F
E
086
.
0
L y
y
y
y
b
D1.2b Stability Bracing of Beams: Highly Ductile Members
Both flanges of beams shall be laterally braced, with a
maximum spacing of Lb = 0.086 ry E / Fy
Note:
For typical SMF beam: ry 2 to 2.5 inches.
and Lb 100 to 125 inches (approx. 8 to 10 ft)
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E3.4b Stability Bracing of Beams
In addition, unless otherwise indicated by testing, beam
braces shall be placed near concentrated forces, changes
in cross section, and other locations where analysis
indicates that a plastic hinge will form during inelastic
deformations of the SMF.
The placement of lateral bracing shall be consistent with
that documented for a prequalified connection
designated in ANSI/AISC 358, or as otherwise determined in
a connection prequalification in accordance with Section K1,
or in a program of qualification testing in accordance with
Section K2.
@alirezasalehin
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ANSI/AISC 358 - Lateral Bracing Requirements for the RBS
Exception for beams with an RBS connection:
Where the beam supports a concrete structural slab
connected between protected zones with welded shear
connectors at 12” o.c. maximum, supplemental top and
bottom flange bracing at reduced section not required.
Otherwise, when a composite concrete floor slab is not
present, supplemental lateral bracing shall be attached
to beam no greater than d/2 beyond end of reduced
section, at the end farthest from the column face.
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E3.4c Stability Bracing at Beam-to-Column
Connections
When the webs of the beams and column are co-planar, and
a column is shown to remain elastic (if ratio above >2.0,
OK!) outside of the panel zone, column flanges at beam-to-
column connections shall require stability bracing only at the
level of the top flanges of the beams.
*
*
2.0?
pc
pb
M
M
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E3.4c Stability Bracing at Beam-to-Column
Connections
When a column cannot be shown to remain elastic outside
of the panel zone, the column flanges shall be laterally
braced at the levels of both the top and bottom beam
flanges. Stability bracing is permitted to be either direct or
indirect.
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Section E3
5. Members
5b. Beam Flanges
5c. Protected Zones
Beams and columns satisfy Section D1.1 for highly
ductile members (unless otherwise qualified).
Beams permitted to be composite with reinforced
concrete slab to resist gravity loads.
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Section E3
5. Members
5b. Beam Flanges
5c. Protected Zones
Drilling of holes or trimming of beam flange width not
permitted unless testing or qualification demonstrates
stable plastic hinges to accommodate required story
drift angle.
Configuration shall be consistent
with ANSI/AISC 358 or
qualification testing.
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Section E3
5. Members
5b. Beam Flanges
5c. Protected Zones
Region at each end of beam subject to inelastic
straining shall be designated as a protected zone.
Extent of protected zone as designated in ANSI/AISC
358 or qualification testing.
In general, for unreinforced
connections, from face of
column to one half beam depth
beyond plastic hinge point.
@alirezasalehin
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E3.6 Connections
Section heavily influenced by 1994 Northridge
Earthquake and subsequent research programs
(SAC-FEMA).
No particular type of connection prescribed; rather,
performance requirements specified.
6a. Demand Critical Welds
6b. Beam-to-Column Connections
6c. Conformance Demonstration
6d. Required Shear Strength
6e. Panel Zone
6f. Continuity Plates
6g. Column Splices
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E3.6a Demand Critical Welds
(1) Groove welds at column splices
(2) Welds at column-to-base plate connections
(3) Complete-joint-penetration groove welds of beam
flanges and beam webs to columns
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(2) Welds at column-to-base plate connections
E3.6a Demand Critical Welds - Exceptions
Where it can be shown that column hinging at, or
near, the base plate is precluded by conditions of
restraint, and in the absence of net tension under
load combinations including the amplified seismic
load, demand critical welds are not required.
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(3) Complete-joint-penetration groove welds of beam
flanges and beam webs to columns
E3.6a Demand Critical Welds - Exceptions
unless otherwise designated by ANSI/AISC 358, or
otherwise determined in a connection prequalification
in accordance with Section K1, or as determined in a
program of qualification testing in accordance with
Section K2..
@alirezasalehin
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E3.6b Beam-to-Column Connections
Beam-to-column connections shall satisfy the following
requirements:
1. The connection shall be capable of
sustaining an interstory drift angle of at
least 0.04 radians.
2. The measured flexural resistance of the
connection, determined at the column
face, shall equal at least 0.80 Mp (nominal)
of the connected beam at an interstory
drift angle of 0.04 radians.
@alirezasalehin
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AISC Seismic Provisions - SMF - Beam-to-Column Connections
E3.6b Requirements
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E3.6d Required Shear Strength
The required shear strength of the connection shall be
determined using load combinations in the applicable
building code, with the following quantity for the
earthquake load effect Emh:
Emh = 2 [ 1.1 Ry Mp ] / Lh (E3-6)
where:
Ry = ratio of the expected yield strength to the minimum
specified yield strength
Mp = nominal plastic flexural strength
Lh = distance between plastic hinge locations
@alirezasalehin
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Lh
(1.2 + 0.2SDS) D + 0.5 L or (0.9-0.2SDS) D
Vu = 2 [ 1.1 Ry Mp ] / Lh + Vgravity
1.1 Ry Mp 1.1 Ry Mp
Vu Vu
Required Shear Strength of Beam-to-Column Connection
Depends on connection type, but
typ. assumed to be db / 2
@alirezasalehin
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E3.6c Conformance Demonstration
Beam-to-column connections used in the SFRS shall satisfy the
requirements of Section E3.6b by one of the following:
(a) Use of SMF connections designed in accordance with
ANSI/AISC 358.
or
(c) Provision of qualifying cyclic test results in
accordance with Section K2. Results of at least two
cyclic connection tests shall be provided and shall be
based on one of the following:
(b) Use of a connection prequalified for SMF in
accordance with Section K1.
@alirezasalehin
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E3.6c Conformance Demonstration
(c) Provision of qualifying cyclic test results in
accordance with Section K2. Results of at least two cyclic
connection tests shall be provided and shall be based on
one of the following:
ii) Tests that are conducted specifically for the project
and are representative of project member sizes, material
strengths, connection configurations, and matching
connection processes, within the limits specified in
Section K2
(i) Tests reported in the research literature or
documented tests performed for other projects that
represent the project conditions, within the limits
specified in Section K2
@alirezasalehin
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Test connection in
accordance with
Chapter K
E3.6c Conformance Demonstration - by Testing
Note: prequalified
connections must
also have been
tested according to
Chapter K before
becoming
prequalified
@alirezasalehin
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Section K2
Cyclic Tests for Qualification of Beam-to-Column
and Link-to-Column Connections
Testing Requirements:
• Test specimens should replicate as closely as practical the
prototype (prototype = actual building)
• Beams and columns in test specimens must be nearly full-
scale representation of prototype members (Section K2.3b):
- depth of test beam ≥ 0.90 depth of prototype beam
- wt. per ft. of test beam ≥ 0.75 wt. per ft. of prototype beam
- depth of test column ≥ 0.90 depth of prototype column
• Sources of inelastic deformation (beam, panel zone,
connection plates, etc) in the test specimen must similar to
prototype.
@alirezasalehin
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Section K2
Testing Requirements (cont):
• Lateral bracing in test specimen should replicate
prototype.
• Connection details used in the test specimen shall
represent the prototype connection details as closely as
possible.
• Welding processes, procedures, electrodes, etc. used for
test specimen must be representative of prototype.
See Section K2 for more specifics and other
requirements.
Additional bracing near loading / reaction points permitted.
@alirezasalehin
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Typical Test Subassemblages
Exterior Subassemblage Interior Subassemblage
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Typical Exterior Subassemblage
Δ
Lbeam
Interstory Drift Angle =
Δ
Lbeam
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Typical Exterior Subassemblage
@alirezasalehin
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Δ
Hcolumn
Typical Interior
Subassemblage
Interstory Drift Angle =
Δ
Hcolumn
@alirezasalehin
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Typical Interior Subassemblage
@alirezasalehin
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Typical Interior Subassemblage (with concrete floor slab)
@alirezasalehin
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Section K2.4
Testing Requirements - Loading History
6 cycles at = 0.00375 rad.
continue at increments of 0.01 rad, with two
cycles of loading at each step
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Section K2.4
Testing Requirements - Loading History
-0.05
-0.04
-0.03
-0.02
-0.01
0
0.01
0.02
0.03
0.04
0.05
0.06
Interstory
Drift
Angle
Acceptance Criteria for SMF Beam-to-Column Connections (Section K2.8):
After completing at least one loading cycle at 0.04 radian, the
measured flexural resistance of the connection, measured at the face
of the column, must be at least 0.80 Mp of the connected beam
@alirezasalehin
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Example of Successful Conformance Demonstration Test
per Section K2:
-40000
-30000
-20000
-10000
0
10000
20000
30000
40000
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Interstory Drift Angle (rad)
Beam
Moment
at
Face
of
Column
(in-kips)
0.8 Mp
- 0.8 Mp
M0.04 0.8 Mp
M0.04 0.8 Mp
@alirezasalehin
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E3.6e Panel Zone of Beam-to-Column Connections
(1) Shear Strength
(2) Panel Zone Thickness
(3) Panel Zone Doubler Plates
@alirezasalehin
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AISC Seismic Provisions - SMF - Panel Zone Requirements
E3.6e (1) Shear Strength
The minimum required shear strength, Ru , of the panel zone shall be
taken as the shear generated in the panel zone when plastic hinges form
in the beams.
To compute panel zone shear.....
Determine moment at beam plastic hinge locations
(1.1 Ry Mp or as specified in ANSI/AISC 358)
Project moment at plastic hinge locations to the face
of the column (based on beam moment gradient)
Compute panel zone shear force.
@alirezasalehin
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Mpr-2
Mpr1
Vbeam-2
Vbeam-1
Beam 1 Beam 2
sh sh
Mf1 Mf2
Mpr = expected moment at plastic hinge = 1.1 Ry Mp or as specified in ANSI/AISC
358
Vbeam = beam shear (see Section E3.6d - beam required shear strength)
sh = distance from face of column to beam plastic hinge location (specified in
ANSI/AISC 358)
Panel Zone Shear Strength (cont’d)
@alirezasalehin
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Mpr-2
Mpr1
Vbeam-2
Vbeam-1
Beam 1 Beam 2
sh sh
Mf1 Mf2
Panel Zone Shear Strength (cont’d)
Mf = moment at column face
Mf = Mpr + Vbeam sh
@alirezasalehin
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Panel Zone Shear Strength (cont)
c
f
b
f
u V
t
d
M
R
Panel Zone Required Shear Strength =
@alirezasalehin
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Panel Zone Design Requirement:
Ru v Rv where v = 1.0
Rv = nominal shear strength, based
on a limit state of shear yielding, as
computed per Section J10.6 of the
AISC Specification
Reminder: Intent of AISC Seismic Provisions is to permit
limited yielding of the panel zone when flexural plastic hinges
have formed in the beams.
@alirezasalehin
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To compute nominal shear strength, Rv, of panel zone,
When Pu 0.75 Py in column:
p
c
b
2
cf
cf
p
c
y
v
t
d
d
t
b
3
1
t
d
F
6
.
0
R
(AISC Spec EQ J10-11)
Where:
dc = column depth
db = beam depth
bcf = column flange width
tcf = column flange thickness
Fy = minimum specified yield stress of column web
tp = thickness of column web including doubler plate
When frame stability, including plastic panel-zone
deformation, is considered in the analysis:
@alirezasalehin
@Seismicisolation
@Seismicisolation

To compute nominal shear strength, Rv, of panel zone:
When Pu > 0.75 Py in column (not recommended):
y
u
p
c
b
2
cf
cf
p
c
y
v
P
P
2
.
1
9
.
1
t
d
d
t
b
3
1
t
d
F
6
.
0
R (AISC Spec EQ J10-12)
@alirezasalehin
@Seismicisolation
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