Fellenius bases de diseño de pilotes de fundación

Bengt H. Fellenius, Dr.Tech., P.Eng.
2475 Rothesay Avenue, Sidney, British Columbia, V8L 2B9
TEL: (778) 426-0775 e-address: <Bengt@Fellenius.net>
Web site: [www.Fellenius.net]
Basics of Design of Piled Foundations
A Course and Seminar
Santa Cruz, Bolivia
April 25, 2013
The primary intent of the course is to demonstrate that deep foundation design is a good deal more than
finding some value of capacity. The course aims to show what data one must pull together and present
processes of analysis and calculations necessary for a design of a specific project. Aspects of negative
skin friction and associated drag load and downdrag are emphasized.
The presentation includes both broad generalities and in-depth details. Aspects of where to install
instrumentation, perform a test, and analyze the test data are addressed. Settlement analysis is of vital
importance to the design of piled foundations, and the course addresses principles of settlement analysis
and provides some of the mechanics of calculating settlement. A few aspects are included of
construction aspects as well as of Limit States Design, LSD (Ultimate Limit States, ULS, and
Serviceability Limit States, SLS, by Canadian terminology and Load and Resistance Factor Design,
LRFD, by US terminology).
To simplify following along the flow of the presentation and taking notes, hand-out course notes are
provided, consisting of black-and-white copies of all Power Points slides, six to a page. Full-size color
copies of the slides are also available on my web site [www.Fellenius.net]. These can be downloaded
from the link [/Bolivia]. Note, the link is hidden and has to be typed into the command line ("command
ribbon").
The slides contain only a minimum of text. For a background and explanation to much of the
presentations, I refer you to my text book "Basics of Foundation Design" also available for downloading
from my web site (the file is called “313 The Red Book_Basics of Foundation Design.pdf”. After
downloading, the book can be viewed and read on-screen or be printed (color or black & white) without
any restriction. The book contains a list of references pertinent to the material presented in the course.
Copies of the referenced papers where I am the author or co-author are available for downloading at my
web site (click on the link "Download Papers").
I will be glad to respond to any e-mail with a question you might wish to put to me.
Sidney April 2013
Bengt H. Fellenius

Basics of Design of Piled Foundations
A Course and Seminar
Bengt H. Fellenius, Dr.Tech., P.Eng.
The course comprises four main lectures leading up to and presenting the essentials of the Unified Method of deep
foundations design for capacity, drag load, settlement, and downdrag for single piles, pile groups, and piled
foundations. The presentations are illustrated with case histories of testing and design analysis including how to
evaluate strain-gage measurements from instrumented pile loading tests and to assess residual load. Settlement
analysis is of vital importance to the design of piled foundations, and the course addresses principles of settlement
analysis and how to calculate settlement of piles and piled foundations. Pertinent aspects of construction
procedures and Load and Resistance Factor Design, LRFD are discussed.
08:00h Brief Background to Basic Principles Applicable to Piled Foundations
Stress distribution and interaction between adjacent foundations; Settlement analysis; Applications of
wick drains to piled foundations.
09:30h Coffee Break
09:45h Analysis of Load Transfer, Capacity, and Response to Load
Load-movement response of foundations; Bearing capacity and load-transfer by beta, alpha, and lambda
methods, and by CPT and CPTu methods; Set-up and relaxation; Residual load; Results of prediction
events.
11:30h The Static Loading Test: Performance, Analysis, and Instrumentation
Methods of testing and basic interpretation of the results. How to analyze results from strain-gage
instrumented piles to arrive at resistance distribution along the pile shaft and the pile toe response.
12.00h LUNCH
13:00h The Static Loading Test: Resumed
Determining pile elastic modulus. The importance of residual load and how to include its effect in the
analysis. Principles of the bi-directional test (the O-cell test) and how to analyze the results of an O-cell
test. Case histories of analyses on results of static loading tests on driven and bored piles.
14:30h Coffee Break
14:50h 4. Piles and Pile Groups — Long-Term Behavior and how we know what we know;
The Unified Design Method.
Important case histories presenting studies that demonstrated the actual long-term response of piles to
load and observed settlement of piles and pile groups. The lessons learnt will be referenced to aspects of
design applying the Unified Method for Design of Piled Foundations considering Capacity, Drag Load,
Settlement, and Downdrag for single piles, pile groups, and piled foundations.
1. Capacity (choice of factor of safety, and rules of LRFD and Limit States Design) and design for
structural strength (including drag load)
2. Settlement of single piles and pile groups due to load directly on the piles and due to influence from
adjacent activity (downdrag)
3. How to combine the various aspects for the design of an actual case with emphasis on foundation
settlement illustrated with examples
17:00h Questions and Discussions; End of Day

1
BASICS OF DESIGN
OF PILED
FOUNDATIONSFOUNDATIONS
Bengt H. Fellenius
1
A short course
Santa Cruz, Bolivia, April 25, 2013
08:00h Brief Background to Basic Principles
Applicable to Piled Foundations
SCHEDULE
09:30h Break
09:45h Analysis of Load Transfer,
Capacity and Response to Load
11.30h The Static Loading Test: Head-down and O-cell Tests
12.00h LUNCH
13.00h The Static Loading Test: Continued
14 00h Case Histories on Pile Analysis Drag Load Downdrag
2
14.00h Case Histories on Pile Analysis, Drag Load, Downdrag,
Pile Groups, Piled Raft, Piled Pad
14.30h Break
14.50h The Unified Method of Design
17:00h Questions and Discussions and End of Day

2
www.Fellenius.net
Bolivia
To Download All
COURSE SLIDES
Power Point Slides
1 - Background Lecture 1.pdf
2 - Analysis Methods Lecture 2.pdf
3 - Static Loading Test Lectures 3a and 3b.pdf
4 - Case Histories and Lectures 4a and 4b.pdf
Design Methods
4

3/24/2013
1
BASICS OF DESIGN
OF PILED
FOUNDATIONS
Bengt H FelleniusBengt H. Fellenius
Background and Basic Principles
Bolivia, April 25, 2013
Some Fundamental Principles
22
Determining the effective stress is
the key to geotechnical analysis
• The not-so-good
method:
hΔ=Δ '' γσ γ’ = buoyant
unit weight
33
)'(' hz Δ∑= γσ
)1(' iwt −−= γγγ
It is much better to determine, separately,
the total stress and the pore pressure.
The effective stress is then the total stress
minus the pore pressure.
)( hΔ∑
44
)( hz Δ∑= γσ
uz −= σσ'
Determining pore pressure
u = γw h
The height of the column of water (h; the head representing the water pressure)
is usually not the distance to the ground surface nor, even, the distance to the
groundwater table. For this reason, the height is usually referred to as the
“phreatic height” or the “piezometric height” to separate it from the depth below
PRESSURE
55
the groundwater table or depth below the ground surface.
The pore pressure distribution is determined by applying the facts that
(1) in stationary conditions, the pore pressure distribution can be assumed to be
linear in each individual soil layer
(2) in pervious soil layers that are “sandwiched” between less pervious layers,
the pore pressure is hydrostatic (that is, the vertical gradient is unity)
SAND
Hydrostatic distribution
CLAY
Non-hydrostatic distribution,
but linear
SAND
Hydrostatic distribution
Artesian phreatic head
GW
DEPTH
Distribution of stress
below a a small load area
0
LB
qqz
×
×=
The 2:1 method
66
)()(
0
zLzB
qqz
+×+
The 2:1-method can only be used for distributions directly under the center
of the footprint of the loaded area. It cannot be used to combine (add)
stresses from adjacent load areas unless they all have the same center. it is
then only applicable under the area with the smallest footprint.

3/24/2013
2
The Boussinesq Method
Derived from calculation of stress from
a point load on the surface of an elastic medium
3
3z
77
2/522
)(2
3
zr
z
Qqz
+
=
π
Newmark’s method for stress
from a loaded area
Newmark (1935) integrated the Boussinesq equation over a finite
area and obtained a relation for the stress under the corner of a
uniformly loaded rectangular area, for example, a footing
CBA
I
+×
88
π4
0
C
Iqqz =×=
2222
22
1
12
nmnm
nmmn
A
+++
++
=
1
2
22
22
++
++
=
nm
nm
B
⎥
⎥
⎦
⎤
⎢
⎢
⎣
⎡
−++
++
= 2222
22
1
12
arctan
nmnm
nmmn
C
m = x/z
n = y/z
x = length of the loaded area
y = width of the loaded area
z = depth to the point under the corner
where the stress is calculated
(1)
• Eq. 1 does not result in correct stress values
near the ground surface. To represent the stress
near the ground surface, Newmark’s integration
applies an additional equation:
π CBA −+×
99
π
π
4
0
CBA
Iqqz
+×
=×=
For where: m2 + n2 + 1 ≤ m2 n2
(2)
Stress distribution below the center
of a square 3 m wide footing
-2
0
)
0 15
0.20
0.25
CTOR,I
Eq. (1)
Eq. (2) Eq. (2)
1010
0 20 40 60 80 100
-6
-4
STRESS (KPa)
DEPTH(m
0.01 0.10 1.00 10.00
0.00
0.05
0.10
0.15
m and n (m = n)
INFLUENCEFAC
Eq. (1)
0
1
2
0 25 50 75 100
STRESS (%)
meters)
Boussinesq
Westergaard
0
1
2
0 25 50 75 100
SETTLEMENT (%)
meters)
Boussinesq
Westergaard
1111Comparison between Boussinesq, Westergaard, and 2:1 distributions
3
4
5
DEPTH(dia
2:1
3
4
5
DEPTH(dia
2:1
0
1
2
0 25 50 75 100
STRESS (%)
eters)
Westergaard
Boussinesq
0
1
2
0 25 50 75 100
SETTLEMENT (%)
meters)
Boussinesq
Westergaard
1212
2
3
4
5
DEPTH(diam
2:1
2
3
4
5
DEPTH(diam
2:1

3/24/2013
3
0
1
2
0 25 50 75 100
STRESS (%)
ameters)
Westergaard
Boussinesq
0
1
2
0 25 50 75 100
SETTLEMENT (%)
ameters)
Boussinesq
Westergaard
1313
3
4
5
DEPTH(dia
2:1 Characteristic
Point; 0.37b
from center
3
4
5
DEPTH(dia
2:1 Characteristic
Point; 0.37b
from center
Below the characteristic point, stresses for flexible and stiff footings are equal
Now, if the settlement distributions are so
similar, why would we persist in using
Boussinesq stress distribution instead of
the much simpler 2:1 distribution?
1414
Because a footing is not alone in this world;
near by, there are other footings, and fills,
and excavation, etc., for example:
The settlement imposed
outside the loaded
foundation is often critical
0
1
2
0 25 50 75 100
SETTLEMENT (%)
meters)
Boussinesq
Outside Point Boussinesq
Center Point
1515
2
3
4
5
DEPTH(diam
Loaded
area
The end result of a
geotechnical design analysis
is
1616
Settlement
Stress-Strain
σ'(KPa)
Δσ

ε
σ
Δ
Δ
=tM
1717
STRAIN (%)
STRESS,σ
Δσ
Δε
Δε
Plotted as Strain-Stress
N(%)N(%)TIO,e
Plotted as Void Ratio vs. Stress
1818
STRESS, σ' (KPa)
STRAIN
STRESS, σ' (KPa)
STRAIN
STRESS (KPa)
VOIDRAT

3/24/2013
4
Stress-strain behavior is non-linear for most soils. The
non-linearity cannot be disregarded when analyzing
compressible soils, such as silts and clays, that is, the
elastic modulus approach is not appropriate for these soils.
Non-linear stress-strain behavior of compressible soils, is
conventionally modeled as follows.
11 '
l
'
l
σσC
1919
where ε = strain induced by increase of effective stress from σ‘0 to σ‘1
Cc = compression index
e0 = void ratio
σ‘0 = original (or initial) effective stress
σ‘1 = final effective stress
CR = Compression Ratio = (MIT)
0
1
0
1
0 '
lg
'
lg
1 σ
σ
σ
σ
ε CR
e
Cc
=
+
=
01 e
C
CR c
+
=
Some use the term "Ccε"
for the "CR", creating
quite a bit of confusion
thereby
In overconsolidated soils (most soils are)
)
'
'
lg
'
'
lg(
1
1 1
00 p
c
p
cr CC
e σ
σ
σ
σ
ε +
+
=
2020
where σ‘p = preconsolidation stress
Ccr = re-compression index
The Janbu Method
The Janbu tangent modulus approach, proposed by Janbu (1963; 1965; 1967; 1998),
and referenced by the Canadian Foundation Engineering Manual, CFEM (1985; 1992),
applies the same basic principles of linear and non-linear stress-strain behavior. The
method applies to all soils, clays as well as sand. By this method, the relation between
stress and strain is a function of two non-dimensional parameters which are unique for a
soil: a stress exponent, j, and a modulus number, m.
2121
Janbu’s general relation is
])
'
'
()
'
'
[(
1 01 j
r
j
rmj σ
σ
σ
σ
ε −=
where: σ‘r = a “reference stress = 100 KPa
j = a stress exponent
m = the modulus number
The Janbu Method
Dense Coarse-Grained Soil j = 1
Cohesive Soil j = 0
1'
ln
1 σ
ε =
'
1
)''(
1
01 σσσε Δ=−=
mm
'
2
1
)''(
2
1
01 σσσε Δ=−=
mm
σ’ in KPa
σ’ in ksf
2222
Cohesive Soil j = 0
Sandy or Silty Soils j = 0.5
0'
ln
σ
ε
m
=
)''(
5
1
01 σσε −=
m
p
m
''(
2
1 σσε −=
σ’ in KPa
σ’ in ksf
There are direct mathematical conversions
between m and the E and Cc-e0
For E given in units of KPa (and ksf), the relation between the
modulus number and the E-modulus is
2323
m = E/100 (KPa)
m = E/2 (ksf)
For Cc-e0, the relation to the modulus number is
cc C
e
C
e
m 00 1
3.2
1
10ln
+
=
+
= And m = 2.3/CR
Typical and Normally Conservative Modulus Numbers
SOIL TYPE MODULUS NUMBER STRESS EXP.
Till, very dense to dense 1,000 — 300 (j=1)
Gravel 400 — 40 (j=0.5)
Sand dense 400 — 250 (j=0.5
compact 250 — 150 _ " _
loose 150 — 100 _ " _
Silt dense 200 — 80 (j=0.5)
compact 80 — 60 _ " _
loose 60 — 40 _ " _
This is where the greater
value of the Janbu
approach versus the MIT
CR-approach comes in.
Clays
Silty clay hard, stiff 60 — 20 (j=0)
stiff, firm 20 — 10 _ " _
Clayey silt soft 10 — 5 _ " _
Soft marine clays
and organic clays 20 — 5 (j=0)
Peat 5 — 1 (j=0)
For clays and silts, the recompression modulus, mr, is often five to twelve times greater than the virgin modulus, m.
This is where the Janbu
approach and the MIT
CR-approach are equal
in practicality.
Reference: Fellenius, B.H., 2012. Basics of foundation design, a text book.
Revised Electronic Edition, [www.Fellenius.net], 385 p.

3/24/2013
5
0.80
1.00
1.20
VoidRatio(--)
m = 12
(CR = 0 18)
p'c
10
15
20
25
Strain(%)
C
1/m
Slope = m = 12
Evaluation of compressibility from oedometer results
2525
0.40
0.60
10 100 1,000 10,000
Stress (KPa) log scale
V
(CR = 0.18)
0
5
10 100 1,000 10,000
Stress (KPa) log scale
p 10p
Cc
Cc = 0.37
e0 = 1.01 p'c
p 2.718p
Void-Ratio vs. Stress and Strain vs. Stress — Same test data
Note, if the "zero"-value -- the e0 -- is off, the Cc-e0 is off (and so is the CR) even if
the Cc is correctly determined. Not so the "m" (if determined from the test results).
Comparison between the Cc/e0 approach
and the Janbu method
0 10
0.15
0.20
0.25
0.30
0.35
PRESSIONINDEX,Cc
Do these values
indicate a
compressible soil, a
medium compressible
soil, a moderately
ibl il
15
20
25
30
35
MODULUSNUMBER,m
2626
Data from a 20 m thick sedimentary deposit
0.00
0.05
0.10
0.40 0.60 0.80 1.00 1.20
VOID RATIO, e0
COMP
compressible soil, or a
non-compressible
soil?0
5
10
0.400.600.801.001.20
VOID RATIO, e0
VIRGIN
The Cc-e0 approach (based on Cc) implies that the compressibility varies by 30± %.
However, the Janbu methods shows it to vary only by 10± %. The modulus number, m,
ranges from 18 through 22; It would be unusual to find a clay with less variation.
Conventional Cc/e0
How many of these
oedometer results indicate
(o) highly compressible clay
(o) compressible clay
( ) di ibl l
20
30
40
50
ODULUSNUMBER,m
0 20 40 60 80 100
WATER CONTENT, wn (%)
Janbu Modulus Number m
The Cc-values converted via the
associated e0-values to modulus
numbers.
2
3
4
5
MPRESSIONINDEX,Cc
2727
(o) medium compressible clay
(o) non-compressible clay?
0
10
0.00 0.50 1.00 1.50 2.00 2.50 3.00
VOID RATIO, e0
VIRGINM
m < 10 ==> Highly compressible
Oedometer test data from
Leroueil et al., 1983
0
1
0.00 1.00 2.00 3.00
VOID RATIO, e0
COM
Stress produces strain
Linear Elastic Deformation (Hooke’s Law)
ε = induced strain in a soil layer
= imposed change of effective stress in the soil layer'σΔ
E
'σ
ε
Δ
=
2828
p g y
E = elastic modulus of the soil layer (Young’s Modulus)
Young’s modulus is the modulus for when lateral expansion is allowed, which may be the case for soil loaded by
a small footing, but not when the load is applied over a large area. In the latter case, the lateral expansion is
constrained (or confined). The constrained modulus, D, is larger than the E-modulus. The constrained modulus
is also called the “oedometer modulus”. For ideally elastic soils, the ratio between D and E is:
ν = Poisson’s ratio
)21()1(
)1(
νν
ν
−+
−
=
E
D
Settlement is due to Immediate Compression,
Consolidation Settlement, and Secondary Compression
Immediate Compression is the compression of the soil grains (soil skeleton) and of any free gas
present in the voids. It is usually assumed to be linearly proportional to the change of stress The
immediate compression is therefore often called 'elastic' compression. It occurs quickly and is
normally small (it is not associated with expulsion of water).
Consolidation (also Primary Consolidation) is volume reduction during the increase in
effective stress occurring from the dissipation of pore pressures (expelling water from the soil
body). In the process, the imposed stress, initially carried by the pore water, is transferred to the
il t t C lid ti i kl i i d il b t l l i fi i d
2929
soil structure. Consolidation occurs quickly in coarse-grained soils, but slowly in fine-grained
soils.
Secondary Compression is a term for compression occurring without an increase of effective
stress. It is triggered by changes of effective stress. It does not usually involve expulsion of
water, but is mainly associated with slow long-term compression of the soil skeleton. Some
compression of the soil structure occurs and it is then associated with some expulsion of water,
but this is so gradual and small that pore pressure change is too small to be noticed. Sometimes,
the term "creep" is used to mean secondary compression, but "creep" should be restricted to
conditions of shear. Secondary compression is usually small, approximately similar in magnitude
to the immediate compression, but may over time add significantly to the total deformation of the
soil over time. Secondary compression can be very large in highly organic soils, such as peat.
Theoretically, seconday compression occurs from the start of the consolidation (effective stress
change), but in practice, it is considered as starting at the end of the consolidation.
On applying load, the soil skeleton compresses and the soil grains
are forced closer to each other reducing the void ratio. The
compression of the soil skeleton occurs more or less immediately in
contrast to the reduction of the void volume which requires
expulsion of water ("consolidation") and can take a long time.
However, in soils containing gas bubbles, the load application
causes the bubbles to compress (and partially to go into solution in
Immediate Compression and Consolidation Settlement
3030
the pore water), which also occurs immediately. Then, as the pore
pressure dissipates during the consolidation process, the gas
bubbles expand which slows down the settlement process. The
"slow-down" is often mistaken for approaching the end of
consolidation. The thereafter observed settlement is then
interpreted as a large secondary compression (addressed later on).

3/24/2013
6
2H
Drainage Layer
Clay Layer
(consolidating)
Drainage Layer
0
1
u
u
S
S
U t
f
t
AVG −==
where UAVG = average degree of consolidation (U)
St = settlement at Time t
Sf = final settlement at full consolidation
ut = average pore pressure at Time t
u0 = initial average pore pressure (on application of the load at Time t = 0)
Basic Relations
UAVG
Consolidation Settlement
3131
v
v
c
H
Tt
2
=
where t = time to obtain a certain degree of consolidation
Tv = a dimensionless time coefficient:
cv = coefficient of consolidation
H = length of the longest drainage path
UAVG (%) 25 50 70 80 90 “100”
Tv 0.05 0.20 0.40 0.57 0.85 ≤1.00
)1(lg1.0 UTv −−−=
HOW TO HANDLE A
MULTILAYERED PROFILE?
c/c
d
"Square" spacing: D = √4/π c/c = 1.13 c/c
"Triangular" spacing: D = √(2√3)/π c/c = 1.05 c/c
Vertical Drains
3232
c/cBasic principle of consolidation process
in the presence of vertical drains
h
h
Ud
D
T
−
−=
1
1
ln]75.0[ln
8
1
hh Ud
D
c
D
t
−
−=
1
1
ln]75.0[ln
8
2
and
h
h
c
D
Tt
2
=
The Kjellman-Barron Formula
Walter Kjellman, inventor of the
very first wick drain, the
Kjellman Wick, a 100 mm wide,
3 mm thick, cardboard drain
that became the prototype for
33
p yp
all subsequent wick drains.
Walter Kjellman, 1950
Important Points
Build-up of Back Pressure
The consolidation process can
be halted if back-pressure is let
to build-up below the
embankment, falsely implying
that the process is completed
3434
Flow in a soil containing pervious lenses, bands, or layers
Theoretically, vertical drains
operate by facilitating horizontal
drainage. However, where
pervious lenses and/or horizontal
seams or bands exist, the water
will drain vertically to the pervious
soil and then to the drain. When
this is at hand, the drain spacing
can be increased significantly.
Pervious seams (silt or
sand) will dry faster than
the main body of clay.
The pervious seams can
be observed in a Shelby
sample during the drying
process, as indicated in
the photos.
3535
p
CPTU soundings with
readings every 10 mm
can also disclose the
presence of sand and silt
seams (if they are thicker
than about 10 mm; which
the illustrated small
seams are not).
How deep do the wick drains have to be installed?
In theory, the drains do not need to go deeper than to where
the applied stress is equal to the preconsolidation stress.
However in practice it is a good rule to always go down to a
3636
However, in practice, it is a good rule to always go down to a
pervious soil layer (aquifer) to ensure downward drainage.
But, if the surcharge is by vacuum treatment or combined
with vacuum treatment, it is better to avoid having the
drains in an aquifer, or they would "suck".

3/24/2013
7
3737
The Kjellman wick, 1942 The Geodrain, 1972
3838
The Geodrain, 1976
Wick drain types
The Burcan Drain, 1978
The Mebra Drain 1984
(a development of the
Castleboard Drain 1979)
3939
0
5
10
15
20
25
30
35
40
0 100 200 300
Pore Pressure (KPa)
Depth(m)
Wick Drains Installed
m)
Settlement at center of a 3.6 m high embankment Bangkok
Airport. Wick drains at c/c 1.5 m were installed to 10 m depth.
PORE
PRESSURE
Enlarged
40
AVERAGE MEASURED
SETTLEMENT
DESIGN CURVE FOR THIS
SURCHARGE (75 KPa)
1.0 m
FINAL HEIGHT
OF FILL
SETTLEMENT(mm)
≈200 days
FILLHEIGHT(m
Calculated
Total
Settlements
Settlement and Horizontal Displacement for the 3.6 m Embankment
WICK DRAINS TO 10 m DEPTH
WICK DRAINS TO 10 m DEPTH
Settlement was monitored in center and at embankment sides and
horizontal displacement was monitored near sides of embankment
Note the
steep slopes
4141
Time from start to end of surcharge placement = 9 months
Observation time after end of surcharge placement = 11 months
1.0 m
2.0 m
WICK
DRAIN
Moh and Lin 2006
Horizontal Displacement versus Settlement at Different Test Locations
OVEMENT(cm)
4242
HORIZONTALMO
SETTLEMENT (cm)

3/24/2013
8
Secondary Compression
1000
log
1 t
t
e
C αα
ε
+
=
The value of the Coefficient of Secondary Compression, Cα, is usually expressed as a
ratio to the consolidation coefficient, Cc, ranging from 0.02 through 0.07 with an average
of about 0.05 (Holtz and Kovacs 1981). For example, in a soft clay with Cc of about 0.3
d f b t it (i d l b f 15) C ld b b t 0 01
4343
and e0 of about unity (i.e., a modulus number of 15), Cα, would be about 0.01.
The key parameter, however, it the t100 value, the time it takes for 100 % of consolidation
(or 90 %, more realistically) to develop. Also when using wick drains, the 100-% should
be the time for vertical drainage, not horizontal.
It is commonly assumed that secondary compression does not start until primary
consolidation is completed; U = 100 %. However, the consensus amongst the experts is
that secondary compression starts as soon as a change of effective stress has been
triggered, i.e., it starts at at 0 % consolidation.
The purpose of calculating stresses is to calculate settlement. The following shows
settlements calculated from the Boussinesq distribution. how stress applied to the
soil from one building affect the settlement of an adjacent existing 'identical'
building loaded the same constructed about 5 years before.
EXISTING
ADJACENT
BUILDING
NEW
BUILDING
WITH SAME
LOAD OVER
FOOTPRINT
AREA
The 2nd building was constructed five years
after the 1st building. The 1st building had
then settled about 80 mm (≈3 inches), which
was OK albeit close to what was felt to be
4444
The soils consist of preconsolidated
(OCR = 2) compressible silt and clay
6.5 m6.5 m 4 m
m
1st
Building
2nd
Building
was OK, albeit close to what was felt to be
acceptable. Did the construction of the 2nd
building add settlement to the 1st, and what
was the settlement of the 2nd building?
(Because the buildings are on raft foundation, the
characteristic point is the most representative point
for the settlement calculations).
The settlement of the first building calculated using UniSettle Version 4
0 2 4 6 8 10
YEARS
SETTLEMENT OVER TIME
4545
0
20
40
60
80
100
120
0 2 4 6 8 10
SETTLEMENT(mm)
2nd Building
constructed
Calculations using Boussinesq distribution can be used to determine how stress
applied to the soil from one building may affect an adjacent existing building
(having the same loading as the new building).
0
5
0 20 40 60 80 100
STRESS (%)
STRESSES
UNDER AREA
BETWEEN THE
TWO BUILDINGS
EXISTING
ADJACENT
BUILDING
NEW
BUILDING
WITH SAME
LOAD OVER
FOOTPRINT
AREA
4646
10
15
20
25
30
DEPTH(m)
STRESSES ADDED
TO THOSE UNDER
THE FOOTPRINT OF
THE ADJACENT
BUILDING
STRESSES
UNDER THE
FOOTPRINT
OT THE
LOADED
BUILDING
TWO BUILDINGS
Calculations by means of UniSettle
The soils consist of preconsolidated
moderately compressible silt and clay
6.5 m6.5 m 4 m
m
Calculations using Boussinesq stress distribution can be used to determine how
stress applied to the soil from one building may affect an adjacent existing building
(having the same loading as the new building).
EXISTING
ADJACENT
BUILDING
NEW
BUILDING
WITH SAME
LOAD OVER
FOOTPRINT
AREA
0
5
10
0 20 40 60 80 100
STRESS (%)
STRESSES
UNDER THE AREA
BETWEEN THE
TWO BUILDINGS
PRECONSOLIDATION
MARGIN (Reducing
with depth)
4747
The soils consist of preconsolidated moderately compressible silt and clay.
The preconsolidation margin reduces with depth.
6.5 m6.5 m 4 m
m
10
15
20
25
30
DEPTH(m)
CENTER
STRESSES
COMBINED
STRESSES
UNDER THE
FOOTPRINT
OF THE
LOADED
BUILDING
STRESSES FROM LOADED
BUILDING CALCULATED
UNDER THE FOOTPRINT
OF THE ADJACENT
BUILDING
Settlement distributions (UniSettle Version 4)
0
5
10
0 20 40 60 80 100 120
SETTLEMENT (mm)
1st
ONLY
Increase due
to 2nd Bldng BOTHSand &
Gravel
Silty
Clay
0
5
10
0 20 40 60 80 100 120
SETTLEMENT (mm)
Of ground
due to 1st
Bldng only
Due to 2nd
Bldng
4848
15
20
25
30
35
DEPTH(m)
1st BUILDING
Soft
Clay 15
20
25
30
35
DEPTH(m)
2nd BUILDING

3/24/2013
9
-83 KPa
105 KPa
34 KPa
85 KPa
105 + 34 + 85 = 224
- 83
141 KPa
110 m
38 m
74 m
MORE ON SETTLEMENT
YEARYEAR
49
Briaud et al. 2007; Fellenius and Ochoa 2008
0
50
100
150
200
250
300
350
400
1936 1946 1956 1966 1976 1986 1996 2006
YEAR
SETTLEMENT(mm).
0
50
100
150
200
250
300
350
400
1 10 100
SETTLEMENT(mm)
1936 1937 1940 1945 1950 1960 1975 2000
LINEAR PLOT
LOWER SCALE
LOGARITHMIC PLOT
UPPER SCALE
1936 1946 1956 1966 1976 1986 1996 2006
0
20
40
60
80
100
120
140
YEAR
WATERDEPTH(m)
132a- 14m
217 - 26m
216a- 39m
115 -153m
209 -159m
111 -161m
501a-180m
912 -206m
114a-261m
618 -267m
606 -301m
501b-365m
132b-442m
114b-480m
1925 1935 1945 1955 1965 1975 1985 1995 2005 2015
SHALLOW WELLS
DEEP WELLS
Water Depths
Measured in
Deep Wells
50
Monument
and Well
Locations
Well head at Burnett School, Baytown, Texas
YEAR
51
0
50
100
150
200
250
300
350
400
1 10 100
YEAR
SETTLEMENT(mm)
1936 1937 1940 1945 1950 1960 1975 2000
DEPTH TO
WATER TABLE
SETTLEMENT
0
25
50
75
100
125
DEPTHTOWATERTABLE(m)
San Jacinto Monument
Settlement and Measured
Depths to Water in the
Wells Plotted Together
1925
The lowering of the pore pressures due to mining of water and subsequent regional
settlement is not unique for Texas. Another such area is Mexico City, for example.
Here is a spectacular 1977 photo from San Joaquin, California.
52
1977
1955
Subsidence at San Joaquin Valley, California
0.0
0.5
1.0
1920 1930 1940 1950 1960 1970 1980 1990 2000 2010
YEAR
ent(m)
I II III IV
5353
1.5
2.0
2.5
Settleme
NEW ORLEANS 1924 - 1978
I. Initial Period of Pumping
II. Increased Pumping
III. Further Increased
IV. Reduced Pumping
Data from Kolb, C.R. and Saucier, RT., 1982
Site Investigation Techniques
The SPT and the CPT/CPTu

3/24/2013
10
The SPT
Example from
Atlantic coast of
South USA
0
5
10
15
0 20 40 60 80 100
SPT N-Indices (bl/0.3m)
0
5
0 10 20 30 40 50
SPT N-Indices (bl/0.3m)
5555
20
25
30
35
40
45
50
DEPTH(m)
East Abutment
10
15
20
25
DEPTH(m)
DETAIL
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H(m)
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H(m)
0
10
20
30
40
0 20 40 60 80 100
N-Index (bl/0.3m)
H(m)
Example from
Atlantic coast of
Canada
5656
40
50
60
70
80
DEPT
40
50
60
70
80
DEPT
40
50
60
70
80
DEPT
SPT for design After problems
arose
Forensics
0
10
20
30
0 20 40 60 80 100
N-Index (bl/0.3m)
m)
With all data points
5757
30
40
50
60
70
80
DEPTH(m
0.010
0.100
1.000
mv
(1/MPa)
30
40
50
60
70
80
90
100
Modulus(MPa)
Direct numerical use
of the SPT N-index
5858
0.001
1 10 100
N60
-Index (bl/0.3m)
0
10
20
0 10 20 30 40 50
N60-Index (bl/0.3m)
(after Terzaghi, Peck, and Mesri 1996
from Burland and Burbidge 1985)
Determining pile Capacity
from SPT-indices
0
5
10
15
0 10 20 30 40
SPT N-Index (bl/0.3m)
(m)
0
5
10
15
0 10 20 30 40
SPT N-Index (bl/0.3m)
(m)
0
5
10
15
0 10 20 30 40
Cone Stress, qt (MPa)
(m)
5959
20
25
30
35
DEPTH(
Estimated required depth
20
25
30
35
DEPTH(
Potentially possible depth
Estimated required depth
1
2
Pile 1 had a much
smaller capacity
than Pile 2!
20
25
30
35
DEPTH(
N (bl/ft)
Pile 1 had a much
smaller capacity
than Pile 2!
2
1
Principles of the CPT and CPTU
The Cone
Penetrometer
606060
Sleeve friction, fs
Pore Pressure
U2 position
Cone Stress, qc
“U2 Position” = pore
pressure measured on
the cone “shoulder”cone shoulder

3/24/2013
11
616161 626262
6363 6464
Continuous cores samples obtained by pushing down a pipe having an
inside plastic tube. On withdrawal and cutting the tube open, the soil
sample is available in a better condition than a sample in a SPT-spoon.
Courtesy of Pinter and Associates, Saskatoon, SK.
0
10
0 10 20 30
INCLINATION ANGLE (°)
(m)
0
10
0 2 4 6 8
RADIAL DEVIATION (m)
(m)
0
10
0.0 0.3 0.5 0.8 1.0
DEPTH DEVIATION (m)
(m)
The CPT sounding rod is never truly vertical, of course.
How much can it be off?
6565
20
30
40
50
ACTUALDEPTH
20
30
40
50
ACTUALDEPTH
20
30
40
50
ACTUALDEPTH
5
10
15
20
25
Y-Direction(m)
20.6 m
PLAN VIEW
"Unfolded"
0
10
20
30
40
50
0 1 2 3 4
DEPTH DEVIATION (m)
EPTH(m)
0
10
20
30
40
50
0 5 10 15 20 25
RADIAL DEVIATION (m)
EPTH(m)
6666
-5
0
-5 0 5 10 15 20 25
X-Direction (m)
Example 2
60
70
80
90
100
DE
60
70
80
90
100
DE
Inclination
plane
X-plane Y-plane

3/24/2013
12
0
5
10
15
0 10 20 30
TH(m)
0
5
10
15
0 100 200
Sleeve Friction (KPa)TH(m)
0
5
10
15
0 100 200 300 400
Pore Pressure (KPa)
TH(m)
0
5
10
15
0.0 1.0 2.0 3.0 4.0
Friction Ratio (%)
TH(m)
CLAY CLAY
CLAY
6767
20
25
30
DEPT
15
20
25
30
DEPT
15
20
25
30
DEPT
20
25
30
DEPT
SILT SILT SILT
SAND SAND SAND
Results of a CPTU sounding
Soil profiling
Applications
6868
The Begemann original profiling chart (Begemann, 1965)
1
10
100
ConeStress,qt(MPa)
4
5
6
7
8
9
10
11
12
Friction Ratio from
0.1 % through 25 %
6969
Profiling chart per Robertson et al. (1986)
0
1 10 100 1,000
Sleeve Friction (KPa)
C
1
2
3 25 %
7070
Profiling chart per Robertson (1990)
1
10
100
ConeStress,qE(MPa)
5 1 = Very Soft Clays, or Sensitive
or Collapsible Soils
2 = Clay and/or Silt
3 = Clayey Silt and/or
Silty Clay
4a = Sandy Silt
4b = Silty Sand
5 = Sand to Sandy Gravel
3
4
7171
0.1
1 10 100 1,000
1 2
The Eslami-Fellenius profiling chart (Eslami 1996; Eslami and Fellenius, 1997)
Example of a CPTU sounding from a river estuary delta (Nakdong River, Pusan, Korea)
0
10
20
30
0 10 20 30
DEPTH(m)
0
10
20
30
0 200 400
DEPTH(m)
0
10
20
30
0 250 500 750 1,000
Pore Pressure (KPa)
DEPTH(m)
0
10
20
30
0 1 2 3 4 5
Friction Ratio (%)
DEPTH(m)
Profile
Mixed
CLAY
7272
The sand layer between 6 m and 8 m depth is potentially liquefiable.
The clay layer is very soft.
The sand below 34 m depth is very dense and dilative, i.e., overconsolidated and
providing sudden large penetration resistance to driven piles and relaxation problems.
30
40
50
30
40
50
30
40
50
30
40
50
SAN
Reduced pore
pressure
(“dilation”)
SAND

3/24/2013
13
1
10
100
oneStress,qE(MPa)
5
3
4
7373
0.1
1 10 100 1,000
Co
1 2
The CPTU data of the Preceding Slide plotted in an Eslami-Fellenius Chart
The CPTU is an excellent and reliable tool for soil
identification, but there is more to geotechnical site
investigation than just establishing the soil type.
And, the CPTU can deliver much more than soil profiling
7474
Liquefaction
7.4 Magnitude Earthquake of August 17, 1999
Kocaeli, Adapazari, Turkey
7575
Photos courtesy of Noel J. Gardner, Ottawa
7676
Photo courtesy of Noel J. Gardner, Ottawa
d
v
v
r
g
a
CSR '
max
65.0
σ
σ
=
CSR = Cyclic Stress Ratio
For earthquake
magnitude of 7.5
An earthquake generates a Cyclic Stress Ratio, CSR
Assessment of liquefaction risk from
results of a CPTU sounding
7777
amax = maximum horizontal acceleration at ground surface (m/s2)
g = gravity constant (m/s2)
σv = total overburden stress (Pa)
σ'v = effective overburden stress (Pa)
rd = stress reduction coefficient for depth (dimensionless)
z = depth below ground surface (m)
CRR
The safety against liquefaction depends on
the Cyclic Resistance Ratio, CRR, determined
from the CPTU data
7878
CSR
CRR
Fs = For earthquake magnitude of 7.5

3/24/2013
14
KPaqfor
q
CRR c
c
5005.0
100
833.0 1
1
<+⎟
⎠
⎞
⎜
⎝
⎛
=
)(045.0 114.0 cq
eCRR =
The following fitted equation represents both equations above
The Cyclic Resistance Ratio, CRR, is expressed in two equations
KPaqKPafor
q
CRR c
c
1605008.0
100
93 1
3
1
<<+⎟
⎠
⎞
⎜
⎝
⎛
=
7979
qc1 = cone stress normalized to depth (i.e., overburden stress)
CNc1 = normalization factor
σr = reference stress = 100 KPa (= atmospheric pressure)
σ'v = effective overburden stress at the depth of the cone
stress considered (KPa)
'11
v
r
ccNcc qCqq
σ
σ
==where
CSR
CRR
Fs =
Determining seismic risk from CPTU sounding
Every plotted point represents an earthquake observation (CSR)
with either no liquefaction of with liquefaction observed
Correlations between CRR-values
calculated from actual earthquakes
versus qc1 values for cases of
liquefaction (solid symbols) and no
liquefaction (open symbols), and
boundary curve (solid line) according
to Robertson and Wride (1998) and
Youd et al. (2001).
The boundary line is the Cyclic
R i t R ti C CRR hi h
All Data; 0 m through 16.0 m
0.4
0.5
0.6
0.7
SR
Robertson and
Wride (1998)
Fines: 35 % 15 %
8080
Resistance Ratio Curve, CRR, which
is also shown as a linear regression
curve for the boundary values. The
two dashed curves show the
boundary curves for sand with fines
contents of 15% and 35%,
respectively (Stark and Olsen 1995).
The original diagram has the cone
stress, qc, divided by atmospheric
pressure to make the number non-
dimensional.
Note, the effect of fines contents has
lately become challenged.
0.0
0.1
0.2
0.3
0 5 10 15 20
Adjusted and Normalized Cone
Stress, qc1 (MPa)
CS
0 m through 6.0 m
0.4
0.5
0.6
0.7
max/g
All Data; 0 m through 16.0 m
0.4
0.5
0.6
0.7
CSR
Separating on two depths and looking at relative seismic
force versus not-normalized cone stress.
Re-analysis of data from Moss et al. (2006)
8181
0.0
0.1
0.2
0.3
0 5 10 15 20
Not Normalized Cone Stress, qc (MPa)
am
A
0.0
0.1
0.2
0.3
0 5 10 15 20
Not Normalized Cone Stress, qc (MPa)
C
B
The 'old' rule that liquefaction risk is small for shallow depth where
the cone stress is ≥5 MPa appears to hold for quake ratio < 0.25.
In the past, liquefaction risk was based on values of the SPT
N-index. Correlations between the CPTU, qc, and the N-index
indicate a ratio between qc and N of about 5. However, that
ratio has a very large range between low and high. It is
questionable how relevant and useful a conversion from an
8282
q
SPT Index value to a cone stress would be for an actual site.
One would be better served pushing a cone in the first place.
Example of determining liquefaction susceptibility before
and after vibratory compaction
0
1
2
3
4
0 5 10 15 20
Cone Stress (MPa)
(m)
0
1
2
3
4
0 50 100 150 200
Pore Pressure (KPa)
H(m)
0
1
2
3
4
0 20 40 60 80
H(m)
0
1
2
3
4
0.0 0.5 1.0
Friction Ratio (%)
H(m)
Sand
PROFILE
Fine sand
to Silty
Sand
8383
5
6
7
8
9
10
DEPTH
5
6
7
8
9
10
DEPTH
5
6
7
8
9
10
DEPTH
5
6
7
8
9
10
DEPTH
Sand
Silty Clay
and Clay
Four CPTU initial (before compaction) soundings at Chek Lap Kok Airport. The heavy
lines in the cone stress, sleeve friction, and friction ratio diagrams are the geometric
averages for each depth of the four soundings.
10
15
ss,qE(MPa)
1 = Very Soft Clays,
Sensitive and/or
Collapsible Soils
2 = Clay and/or Silt
3 = Clayey Silt and/or
Silty Clay
4a = Sandy Silt and/or
Silt
5
Soil chart
8484
0
5
0 20 40 60 80 100
ConeStres
4b = Fine Sand and/or
Silty Sand
5 = Sand to Sandy Gravel
4b
4a
3
2
1

3/24/2013
15
0
1
2
3
4
5
0 5 10 15
Cone Stress (MPa)
TH(m)
0
1
2
3
4
5
0 10 20 30 40 50
TH(m)
0
1
2
3
4
5
0 50 100 150 200
Pore Pressure (KPa)
TH(m)
0
1
2
3
4
5
0.0 0.1 0.2 0.3 0.4 0.5
Friction Ratio (%)
TH(m)
7 Days
7 Days
Before
8585
6
7
8
9
10
DEPT
6
7
8
9
10
DEP
6
7
8
9
10
DEPT
6
7
8
9
10
DEPT
7 DaysBefore
Before
Geometric average values of cone stress, sleeve friction, and friction ratios and
measured pore pressures from CPTU soundings at Chek Lap Kok Airport before
and seven days after the vibratory compaction.
Fs versus depth
0
1
2
3
4
5
0.00 1.00 2.00 3.00 4.00 5.00
Factor of Safety, Fs (--)
PTH(m)
Before
Compaction
7 Days after
CSR
CRR
Fs =
8686
Factor of safety against liquefaction before and after vibratory compaction
6
7
8
9
10
DEP
compaction
CPT and CPTU Methods
for Calculating the Ultimate
Resistance (Capacity) of a Pile
Schmertmann and Nottingham (1975 and 1978)
8787
Meyerhof (1976)
deRuiter and Beringen (1979)
LCPC, Bustamante and Gianeselli (1982 )
Eslami and Fellenius (1997 )
ICP, Jardine, Chow, Overy, and Standing (2005)
But we will save those methods for later
Vibrations from Pile Driving
v =
433 Eh
ZP
M g h
r
=
433 Eh
ZP
M g h
x2
+ z2
V = vertical component of the ground vibration, m/s
Eh = hammer efficiency coefficient
ZP il i d N /
88
ZP = pile impedance, Ns/m
M = hammer (ram) mass, N
G = acceleration, m/s2
H = hammer height-of-fall, m, taken as the equivalent
height-of-fall that corresponds to the kinetic energy
at impact
z = pile penetration depth, m
x = horizontal distance at the ground surface from pile
to observation point, m
Most ground vibrations are
generated from the pile toe
6
8
10
12
14
16
18
20
bration Velocity, v0 (mm/s)
89
0
2
4
0 5 10 15 20 25 30 35 40 45 50
Distance to pile toe, r (m)
Vi
Vibrations from driving a long toe bearing pile: measured compared to calculated

3/24/2013
1
BASICS OF DESIGN
OF PILED
FOUNDATIONS
Bengt H Fellenius
1
Bengt H. Fellenius
Load Transfer and Capacity of Piles
Bolivia, April 25, 2013 22Driving closed-toe pipe piles into fine sand about 2.5 m above the groundwater table
33
Driving 12-inch precast concrete pile into clay for Sidbec in 1974
Head measured in aquifer
below the clay layer
44Svärta River 1969
GW
What really is Capacity?
For piles, capacity is
what we determine in
55
— define from —
a loading test
?
e.g.: The Offset Limit Load (Davisson, 1972)
Do you agree that this point
on the curve represents the
capacity of the pile?
Qu
Qu
66
Rs
Rt

3/24/2013
2
γγ NbNqNcr qcu '5.0'' ++=
and for Footings?
The Bearing Capacity Formula
where ru = ultimate unit resistance of the footing
c’ = effective cohesion intercept
B = footing width
’ b d ff ti t t th f d ti l l
77
q’ = overburden effective stress at the foundation level
γ‘ = average effective unit weight of the soil below the foundation
Nc, Nq, Nγ = non-dimensional bearing capacity factors
The main factor is the
“Nq”
Nq
88
Nq
But what is the reality?
φ
Results of static loading tests on 0.25 m to 0.75 m square
footings in well graded sand (Data from Ismael, 1985)
400
500
600
700
D(KN)
1.00 m
0.75 m
0.50 m
0.25 m
1,000
1,200
1,400
1,600
1,800
2,000
SS(KPa)
Normalized
99
0
100
200
300
0 10 20 30 40 50
SETTLEMENT (mm)
LOAD
MOVEMENT
0
200
400
600
800
,
0 5 10 15 20
MOVEMENT/WIDTH (%)
STRES
1.00 m
0.75 m
0.50 m
0.25 m
Normalized
0
2
4
0 5 10 15 20
0
2
4
0 100 200 300 400
Sleeve Friction, fs (KPa)
0
2
4
0 20 40 60 80
Pore Pressure (KPa)
0
2
4
0 1 2 3 4 5
Friction Ratio, fR (%)
SAND
CPTU PROFILE
Load-Movement for Five Footings on Sand
at Texas A&M University Experimental Site.
J-L. Briaud and R.M. Gibbens, 1994, ASCE GSP 41,
10
6
8
10
12
14
16
DEPTH(m)
6
8
10
12
14
16
DEPTH(m)
6
8
10
12
14
16
DEPTH(m)
6
8
10
12
14
16
DEPTH(m)
SANDY CLAYEY
SILT
Eslami- Robertson
Fellenius
As before the data will tell us
more, if we divide the load with
the footing area (to get stress)
and divide the movement with
the footing width, as follows.
Load-Movement of Four Footings on Sand
Texas A&M University Experimental Site
ASCE GSP 41, J-L Briaud and R.M.
Gibbens 1994
8,000
10,000
12,000
N)
3.0 m
3.0 m
1,400
1,600
1,800
2,000
KPa)
Texas A&M
Settlement Prediction Seminar
11
0
2,000
4,000
6,000
,
0 50 100 150 200
MOVEMENT ( mm )
LOAD(KN
1.5 m
1.0 m
2.5 m
0
200
400
600
800
1,000
1,200
0 5 10 15 20
MOVEMENT / WIDTH (%)
STRESS(
Load-Movement of Four Footings on Sand
Texas A&M University Experimental Site
ASCE GSP 41, J-L Briaud and R.M.
Gibbens 1994
8,000
10,000
12,000
N)
3.0 m
3.0 m
1,600
2,000
)
e
Q
Q
⎟⎟
⎠
⎞
⎜⎜
⎝
⎛
=
2
1
2
1
δ
δ
e = 0.4
q-z curve:
We can also borrow from pile
analysis (Pile toe response) and
apply a q-z function to the stress-
movement data. The "Ratio" function
is applied here.
Texas A&M
12
0
2,000
4,000
6,000
,
0 50 100 150 200
MOVEMENT ( mm )
LOAD(KN
1.5 m
1.0 m
2.5 m
0
400
800
1,200
0 5 10 15 20
MOVEMENT/WIDTH, δ (%)
STRESS,σ(KPa)

3/24/2013
3
Lehane et al. 2008
200
250
300
350
400
450
500
OAD(KN)
1.0 m 1.5 m
1.0 m
200
300
400
500
RESS(KPa)
1.0 m
13
Lehane, B.M., Doherty, J.P., and Schneider, J.A., 2008.
Settlement prediction for footings on sand. Conference
on Deformational Characteristics of Geomaterials. S.E.
Burns, P.W. Mayne, and J.C. Santamarina (Editors),
Atlanta, September 22-24, 2008, Vol. 1, pp.133-150.
0
50
100
150
0 10 20 30 40 50
MOVEMENT (mm)
L
0
100
0 1 2 3 4 5 6 7 8
MOVEMENT / WIDTH (%)
STR
Footing, 1.5 m
Footing 1.0 m
Footing 1.0 m
Six footings on gravel
Caisson under air
pressure to control
water level.
GW
////////// //////////
14 m
16 m
6,000
8,000
10,000
12,000
14,000
TRESS(KPa)
0.3 x 0.3
14
Kusakabe, O., Maeda, Y., and Ohuchi, M., 1992. Large-scale loading tests of shallow footings in
pneumatic caisson. ASCE Journal of Geotechnical Engineering, 118(11) 1681-1695.
"SCORIA" = Sandy GRAVEL, trace fines.
An "interlocked" and highly
overconsolidated volcanic soil.
e0 = 1.2, wn = 40 %, ρ = 1,800 kg/m
3
`
`W
Footing test
!?
0
2,000
4,000
0 5 10 15 20 25 30 35 40
NORMALIZED MOVEMENT (%)
ST
0.3 x 0.3
0.4 x 0.4
0.7 X 0.7
1.3 X 1.3
0.4 X 1.2
0.4 X 2.0
8,000
10,000
12,000
14,000
ESS(KPa)
Considering the "Preloading"/"Reloading"/"Prestress" Effect
15
0
2,000
4,000
6,000
0 5 10 15 20 25 30 35 40
NORMALIZED MOVEMENT (%)
STRE
0.3 x 0.3
0.4 x 0.4
0.7 X 0.7
1.3 X 1.3
0.4 X 1.2
0.4 X 2.0
Data from Kusabe et al. 1992
Plate loading tests on 0.55 m x 0.65 m and 1.10 m x 1.30 m
rectangular slabs in silty sand at Kolbyttemon, Sweden
1,500
2,000
(KPa)
TREND
1 1m x 1 3m
16
Fellenius (2011). Data from Bergdahl, U., Hult, G., and Ottosson, E. (1984)
0
500
1,000
0 1 2 3 4 5 6 7 8 9 10
MOVEMENT (%)
STRESS
0.55m x 0.65m
1.1m x 1.3m
Ultimate Shaft Resistance
rs, Rs
Ultimate Shaft Resistance
is a reality
1717
Ultimate Toe Resistance
does not exist other than
as a definition of load at a
certain movement
rt, Rt
does not exist other than
as a definition of load at a
certain movement
is not
50
100
150
200
AGESHAFTSHEAR
(KPa)
O-cell to GL3
GL3 to GL1
Pile D2000
2,000
3,000
4,000
RAGESTRESSAND
SHEAR(KPa)
Toe Resistance
Pile D2000
Shaft and toe resistances from full-scale static loading tests
on a 2,000 m diameter, 85 m long bored pile in silty clay
Shaft Resistance Toe Resistance
1818
0
50
0 20 40 60 80 100
MOVEMENT (mm)
AVER
0
1,000
0 20 40 60 80 100
MOVEMENT (mm)
AVER
S
Shaft resistances
(repeated for reference)
The above curve shows the shape of the
load-movement every toe resistance.
"Ultimate Toe Resistance" does not exist!
A pile toe reacts to load by a stiffness
response and failure does not occur
however much the pile toe is moved
down.

3/24/2013
4
• Pile capacity is the combined effect of
shaft resistance and toe resistance.
• Shaft resistance is governed by shear
strength, which has an ultimate value.
That is, shaft capacity is reality.
• In contrast, toe resistance is governed by
1919
In contrast, toe resistance is governed by
compression, which does not have an
ultimate value. As the load is increased,
a larger and larger soil volume is
stressed to a level that produces
significant compression, but no specific
failure or peak value: Toe capacity is a
delusion.
Analysis Methods for Determining
Shaft Resistance, rs
The Total Stress Method
The Lambda Method
Th SPT M th d
2020
The SPT Method
The CPT and CPTU Methods
The Pressuremeter Method
The Beta Method
where rs = unit shaft resistance
τu = undrained shear strength
α = reduction coefficient for τu > ≈100 KPa
[ ]uusr αττ ==
Piles in Clay
Total Stress Method
"Alpha analysis"
2121
The undrained shear strength can be obtained from unconfined
compression tests, field vane shear tests, or, to be fancy, from
consolidated, undrained triaxial tests. Or, better, back-calculated from
the results of instrumented static loading tests. However, if those tests
indicate that the unit shaft resistance is constant with depth in a
homogeneous soil, don’t trust the analysis!
2222
Clay adhering to extracted piles
Photo courtesy of K.R. Massarsch
The Lambda Method
Vijayvergia and Focht (1972)
)2'( mms cr += σλ
where rm = mean shaft resistance along the pile
λ = the ‘lambda’ correlation coefficient
σ’m = mean overburden effective stress
cm = mean undrained shear strength
Piles in Clay
2323
Approximate Values of λ
Embedment λ
(Feet) (m) (-)
0 0 0.50
10 3 0.36
25 7 0.27
50 15 0.22
75 23 0.17
100 30 0.15
200 60 0.12
The Lambda method was developed for long piles in clay deposits (offshore conditions)
{ } 'tan')/2()()(lg87.0)(016.02.28.0 2.042.0
δσ zts bhOCRSOCRr −
−+=
OCR = overconsolidation ratio
St = sensitivity
Piles in Clay
A method from fitting a variety of parameters to results from static loading tests
2424
ICP (Imperial College Pile method)
Jardine, Chow, Overy, and Standing (2005 )
h = height of point above pile toe ; h ≤ 4b
b = pile diameter
δ’ = interface friction angle

3/24/2013
5
The SPT Method
Meyerhof (1976)
Rs = n N As D
where Rs = ultimate shaft resistance
n = a coefficient
N = average N-index along the pile shaft (taken as a pure number)
Piles in Sand
2525
g g p ( p )
As = unit shaft area; circumferential area
D = embedment depth
n = 2·103 for driven piles and 1·103 for bored piles (N/m3)
[English units: 0.02 for driven piles and 0.01 for bored piles (t/ft3)]
For unit toe resistance, Meyerhof's method applies the N-index at the pile toe
times a toe coefficient = 400·103 for driven piles and 120·103 for bored piles (N/m3)
[English units: 4 for driven piles and 1 for bored piles (t/ft3)]
CPT and CPTU Methods
for Calculating the Ultimate
Resistance (Capacity) of a Pile
Schmertmann and Nottingham (1975 and 1978)
2626
deRuiter and Beringen (1979)
Meyerhof (1976)
LCPC, Bustamante and Gianeselli (1982 )
ICP, Jardine, Chow, Overy, and Standing (2005)
Eslami and Fellenius (1997 )
caOCRt qCr =
The CPT and CPTU Methods
where rt = pile unit toe resistance (<15 MPa)
COCR = correlation coefficient governed by the
Schmertmann and Nottingham
(1975 and 1978)
CLAY and SAND
SAND (alternative)ccs qKr =
sfs fKr =
2727
overconsolidation ratio, OCR, of the soil
qca = arithmetic average of qc in an influence zone*)
Kf = a coefficient depends on pile shape and material,
cone type, and embedment ratio. In sand, the
coefficient ranges from 0.8 through 2.0, and, in
clay, it ranges from 0.2 through 1.25.
Kc = a dimensionless coefficient; a function of the pile
type, ranging from 0.8 % through 1.8 %
qc = cone resistance (total; uncorrected for pore
pressure on cone shoulder)
*) The Influence zone is 8b above and 4b below pile toe
2828
Filtering of qc-values and determining pile toe
resistance (Schmertmann method)
deRuiter and Beringen
(1979)
uct SNr =
us Sr α=
Means turning the CPT-
method into the Total
St th d
2929
where rt = pile unit toe resistance
Nc = conventional bearing capacity factor
Su = undrained shear strength — — — — —>
NK = a dimensionless coefficient, ranging from 15
through 20, reflecting local experience
α = adhesion factor equal to 1.0 and 0.5 for
normally consolidated and overconsolidated
clays, respectively
An upper limit of 15 MPa is imposed for rt
k
c
u
N
q
S =
Stress method
LCPC
Bustamante and Gianeselli (1982 )
cs qKr =
cat qCr =
3030
C = toe coefficient ranging from 0.40 through 0.55
qca = cone stress averaged in a zone 1.5 b above and
1.5 b below the pile toe plus filtering
rt = pile unit toe resistance < 15 KPa, <35 KPa, or <120 KPa,
depending on soil type, pile type, and pile installation method
K = a dimensionless coefficient; a function of pile type, ranging
from 0.5 % through 3.0 % (Compare: Schmertmann proposes 0.8 %
through 1.8 %)

3/24/2013
6
Soil Type Cone Stress Bored Piles Driven Piles Maximum rt
CLCPC CLCPC
(MPa) (- - -) (- - -) (MPa)
CLAY - - qc < 1 0.04 0.50 15
Coefficients and Limits of Unit Toe Resistance in the LCPC Method Quoted from the CFEM (1992)
3131
c
1 < qc < 5 0.35 0.45 15
5 < qc - - - 0.45 0.55 15
SAND - - - qc < 15 0.40 0.50 15
12 < qc - - - 0.30 0.40 15
Soil Type Cone Stress Concrete Piles Steel Piles Maximum rs
(MPa) & Bored Piles
KLCPC KLCPC J
(- - -) (- - -) (KPa)
CLAY - - qc < 1 0.011 (1/90) 0.033 (=1/30) 15
1 5 0 025 (1/40) 0 011 ( 1/80) 35
Coefficients and Limits of Unit Shaft Resistance in the LCPC Method Quoted from the CFEM (1992)
3232
1 < qc < 5 0.025 (1/40) 0.011 (=1/80) 35
5 < qc - - - 0.017 (1/60) 0.008 (=1/120) 35
SAND - - - qc < 5 0.017 (1/60) 0.008 (=1/120) 35
5 < qc < 12 0.010 (1/100) 0.005 (=1/200) 80
12 < qc - - - 0.007 (1/150) 0.005 (=1/200) 120
The values in the parentheses are the inverse of the KLCPC-coefficient
ca
c
t q
d
b
r )5.01( −=
cJs qKr =
σ ' b
ICP (Imperial College Pile method)
Jardine, Chow, Overy, and Standing (2005 )
3333
δσ
σ
σ
tan)')()
'
(0145.0( 38.013.0
m
tr
z
cJ
h
b
qK Δ+=
b
q
qq
rz
rzccm
c
01.0
)]
'
(10216.1)'(00125.00203.0(2[' 1
2
65.0 −−−
∗−+=Δ
σσ
σσσ
Egtt qCr =
Eslami and Fellenius
(1997 )
Ess qCr =
rt = pile unit toe resistance
Ct = toe correlation coefficient (toe adjustment factor)—equal
to unity in most cases
Shaft Correlation Coefficient
Soil Type*) Cs
Soft sensitive soils 8 0 %
b
Ct
3
1
=
b
Ct
12
=
b in metre
b in inch
3434
qEg = geometric average of the cone point resistance over the
influence*) zone after correction for pore
pressure on shoulder and adjustment to
“effective” stress
rs = pile unit shaft resistance
Cs = shaft correlation coefficient, which is a function of soil
type determined from the soil profiling chart
qE = cone point resistance after correction for pore pressure
on the cone shoulder and adjustment to “effective” stress
*) The Influence zone is 8b above and 4b below pile toe
Soft sensitive soils 8.0 %
Clay 5.0 %
Stiff clay and
Clay and silt mixture 2.5 %
Sandy silt and silt 1.5 %
Fine Sand and silty Sand 1.0 %
Sand to sandy gravel 0.4 %
*) determined directly from the
CPTU soil profiling
Unit shaft resistance as a function of cone stress, qc in Sand
according to the LCPC method and compared to the Eslami-
Fellenius method
100
120
140
ce,rs(KPa)
Sandy Silt to silty Sand to sandy Gravel
Concrete
Range for the Eslami Fellenius method
3535
0
20
40
60
80
0 5 10 15 20 25 30 35 40
Cone Stress, qc (MPa)
UnitShaftResistan
piles
Steel
piles
PILES IN SAND
Cone Stress, qc and qt (MPa)
Pile Capacity or, rather,
Load-Transfer follows
principles of effective stress
3636
principles of effective stress
and is best analyzed using the
Beta method

3/24/2013
7
the Beta method
Unit Shaft
Resistance, rs
zsr 'βσ=
where c‘ = effective cohesion intercept
β = Bjerrum-Burland coefficient
σ'z = effective overburden stress
Effective Stress Analysis (Beta-analysis as opposed to Alpha analysis)
3737
dzcAdzrAR zssss )''( βσ+∫=∫=Total Shaft
Resistance, Rs
where As = circumferential area of the pile at Depth z
(surface area over a unit length of the pile)
Shaft Resistance — in Sand and in Clay
KMr ''tan σφ=
vsr 'σβ=
3838
M = tan δ’ / tan φ’
Ks = earth stress ratio = σ’h / σ’v
σ‘v = effective overburden stress
vss KMr tan σφ=
Approximate Range of Beta-coefficients
SOIL TYPE Phi Beta
Clay 25 - 30 0.20 - 0.35
Silt 28 - 34 0.25 - 0.50
Sand 32 - 40 0.30 - 0.90
Gravel 35 - 45 0.35 - 0.80
0.05 - 0.80 !
3939
Gravel 35 45 0.35 0.80
These ranges are typical values found in some cases. In any given case,
actual values may deviate considerably from those in the table.
Practice is to apply different values to driven as opposed to bored piles, but ....
2.0
3.0
4.0
5.0
6.0
coefficientinsand
G
Trend line
4040
0.0
1.0
0 5 10 15 20 25 30
LENGTH IN SOIL (m)
ß-c
HK GEO (2005)CFEM (1992)
Gregersen
et al. 1973
Beta-coefficient versus embedment length for piles in sand (Data from
Rollins et al. 2005). Ranges suggested by CFEM (1993), Gregersen et al
1973, and Hong Kong Geo (2005) have been added.
1.00
1.50
2.00
2.50
OEFFICIENTINSAND
Concrete piles
Open-toe pipe piles
Closed-toe pipe piles
Gregersen
4141
0.00
0.50
0 50 100 150 200 250 300 350
AVERAGE EFFECTIVE STRESS, σ'z (KPa)
ß-CO
et al. 1973
Beta-coefficient versus average σ’ for piles in sand. (Data from Clausen et al. 2005).
1.00
1.50
2.00
2.50
FICIENTINSAND
Concrete piles
Open-toe pipe piles
Closed-toe pipe piles
4242
0.00
0.50
0.0 0.2 0.4 0.6 0.8 1.0 1.2
AVERAGE DENSITY INDEX, I D
ß-COEF
Beta-coefficient versus average ID for piles in sand. (Data from Karlsrud et al. 2005).

3/24/2013
8
0.20
0.30
0.40
0.50
0.60coefficientinclay
Norway
Japan
Thailand
Vancouver
Alberta
4343
0.00
0.10
0 20 40 60 80
PLASTICITY INDEX, I P
ß-c
Beta-coefficient versus average IP for piles in clay. (Data from Karlsrud et al. 2005
with values added from five case histories).
c
CC
I
CK
r
vD
C eC
e
K
φβ σ
σ
tan
'
ln
100
1
0
24
30
2
⎟⎟
⎠
⎞
⎜⎜
⎝
⎛
−
−
=
Where K = coefficient of earth stress at rest
I = density index (“relative density”)
The Beta-coefficient has a certain appeal to the academia it seems.
This is what is proposed in a recent issue of the ASCE Journal.
44
ID = density index ( relative density )
σ’v = effective overburden stress
σr = reference stress = 100 KPa
Φ = triaxial-compression critical-state
friction angle
C1 = a coefficient: 0.6< C1 <0.7
C2 = a constant = 0.2
Unit Toe
Resistance, rt
where Nt = toe “bearing capacity” coefficient
D = depth to pile toe
σ'z=D = effective overburden stress at the pile toe
Dztt Nr == 'σ
Toe Resistance
4545
Total Toe
Resistance, Rt
where At = toe area (normally, the cross sectional area of the pile)
Dzttttt NArAR === 'σ
Approximate Range of Nt-coefficients
SOIL TYPE Phi Nt
Clay 25 - 30 3 - 30
Silt 28 - 34 20 - 40
Sand 32 - 40 30 - 150
Gravel 35 - 45 60 - 300
4646
The Toe Resistance, Rt, while not really an “ultimate” resistance, is
usually considered as such in design. However, toe resistance should be
thought of as that mobilized in a static loading test at the maximum
acceptable movement usually considered applicable to a piled foundation.
Also the toe resistance appears to have certain qualities
intriguing to the academia. This is what is proposed in
the same recent issue of the ASCE Journal.
DDccD ICC
r
hICCC
r
IC
toeu eCeCr 876542
)
'
()(
21,
−−+−
=
σ
σ
σ φφ
Where ru, toe = ultimate toe resistance for a pile head movement
equal to 10 % of the pile diameter
ID = density index (“relative density”)
!!!
47
D y ( y )
σ’h = effective horizontal stress (= σ’v/K0?)
Φ = triaxial-compression critical-state friction angle
Total Resistance (“Capacity”)
tsult RRQ +=
suzsuz RQdzAQQ −=∫−= '
σβ
0
5
10
0 500 1000 1500 2000
LOAD
H
Qult/ Rult
4848
15
20
25
DEPTH
Rt
Rs
Effective stress — Beta — analysis is the
method closest to the real response of a
pile to an imposed load

3/24/2013
9
0
1
2
0 50 100 150
UNIT SHAFT RESISTANCE (KPa)
0
1
2
0 100 200 300 400 500 600 700 800
TOTAL SHAFT RESISTANCE (KN)
Pile C
CPT-3
Calculations of unit and total shaft resistances for a pile driven into a
saprolite (residual soil) in Porto, Portugal. The soil can be classified both
as a clay type and sand type.
Shaft resistance by CPT-methods
4949
3
4
5
6
DEPTH(m)
Dutch
Sand
Meyerhof
Sand
LCPC
Sand
LCPC
Clay
Schmertmann
Clay
Eslami-
Fellenius
Schmertmann
Sand
Dutch
Clay
Tumay
Clay
a
3
4
5
6
DEPTH(m)
Effective
Stress
Beta = 1.00
Dutch
Sand
Meyerhof
Sand
LCPC
Clay &
Sand
Schmertmann
Clay
Eslami-
Fellenius
Schmertmann
Sand
Dutch
Clay
Tumay
Clayb
0
1
2
0 200 400 600 800 1,000 1,200 1,400 1,600 1,800 2,000
CALCULATED PILE RESISTANCE (KN)
Tumay
Clay
Eslami-
Fellenius
Schmertmann
Clay
Dutch
Clay
LCPC
Dutch
Sand
Meyerhof
Sand
Pile C
CPT-3
Total resistance by CPT-methods
5050
3
4
5
6
DEPTH(m)
Schmertmann
Sand
LCPC
Sand
LCPC
Clay
a
Let’s look at a few case studies
Annacis/Lulu Island
Tests, Vancouver,
BC
by UBC 1985
5151
Static loading tests
on three 324 mm
diameter pipe piles
driven to depths of
14 m, 17 m, and 31 m
into the Fraser River
deltaic soils
0
5
10
15
20
0 5 10 15
PTH(m)
0
5
10
15
20
0 100 200
TH(m)
0
5
10
15
20
0 500 1,000
Pore Pressure (KPa)
PTH(m)
0
5
10
15
20
0 1 2 3 4 5
Friction Ratio (%)
PTH(m)
PILES
1 2 3 4
PROFILE
Eslami-Fellenius Robertson
CLAY CLAY
SANDSAND
SAND
GRAVEL
& SAND
CPT and CPTU analysis for capacity
5252
25
30
35
40
DEP
25
30
35
40
DEP
25
30
35
40
DEP
25
30
35
40
DEP
CLAY
and
Silty
CLAY
CLAY
and
Silty
CLAY
Annacis/Lulu Island Tests
by UBC 1985
The results of the
load-movement
curves from all three
tests combined in
600
800
1,000
1,200
OAD(KN)
Depth 16.8 m
Set-up Time
85 days
Depth 31.1 m
Set-up Time
38 days
5353
Data from Lulu Island Tests
by UBC 1985
tests combined in
one graph. (With offset
limit lines and maximum
load in the tests).
0
200
400
0 10 20 30 40
MOVEMENT (mm)
LO
Depth 13.7 m
Set-up Time
197 days
Results of CPT and CPTU analysis compared to
capacity from the static loading tests
0
5
10
0 500 1,000 1,500 2,000
SHAFT RESISTANCE (KN)
Eslami-Fellenius
Dutch
LCPC
Schmertmann
UniPile eff.stress
ß = 0 15
0
5
10
0 500 1,000 1,500 2,000
SHAFT and TOE RESISTANCEs (KN)
Eslami-Fellenius
Dutch
LCPC
Schmertmann
UniPile eff. stress
Pile static tests
ß = 0.15
5454
“UniPile eff.stress” is effective stress analysis matched to results of static tests
15
20
25
30
35
DEPTH(m)
ß = 0.15
ß = 0.20
ß = 0.15
15
20
25
30
35
DEPTH(m)
ß 0.15
Nt = 7
ß = 0.20
Nt = 25
ß = 0.15
Nt = 3
Test too soon
after EOID

3/24/2013
10
150
a)
O-cell to GL3 GL3 to GL2 GL2 to GL1
O-cell to GL2 O-cell to GL1
Sunrise City Project, HoChiMinh City, Vietnam
1,800 mm diameter bored piles constructed to 70 m depth
Unit shaft resistances versus measured downward movement at
depths of ≈50 m
150
Pa)
O-cell to GL4 GL4 to GL3 GL3 to GL2
O-cell to GL3 O-cell to GL2 O-cell to GL1
SHAFT RESISTANCE
HoChiMinh
Ha Noi
Cai Mep Port
55
0
25
50
75
100
125
0 1 2 3 4 5 6 7 8 9 10
MOVEMENT (mm)
UNITSHAFTRESISTANCE(KPa
TBP-1
Next reading was at 56 mm
ß = 0.14
0
25
50
75
100
125
0 1 2 3 4 5 6 7 8 9 10
MOVEMENT (mm)
UNITSHAFTRESISTANCE(KP
TBP-2
ß = 0.13
Next reading was at 35 mm
No records were obtained during the sudden movement occurring at about 5 mm
0
500
1,000
1,500
2,000
2,500
0 50 100 150 200
MOVEMENT (mm)
UNITRESISTANCE(KPa)
TBP-1
Unit Toe Resistance
Unit Shaft Resistances
10% of diameter
TOE RESISTANCE
56
0
500
1,000
1,500
2,000
2,500
0 50 100 150 200
MOVEMENT (mm)
UNITRESISTANCE(KPa)
TBP-2
TBP-1
Unit Toe Resistance
Unit Shaft Resistances
The stiffness of the toe stress-
movement is unusually soft for a
dense sand and typical of a pile
having a layer of debris at the bottom
of the shaft when the concrete was
placed. A pile a few metre to the side
was constructed using the same
method and equipped with a coring
tube. Coring through this pile toe into
the soil two weeks after construction
revealed presence of about 30 mm of
soft material between the pile and the
soil.
Core from the pile toe and into the soil below
57
Bridge over Panama Canal, Paraiso Reach, Republic of Panama
O-cell test on a 2.0 m (80 inches) diameter, 30 m (100 ft) deep shaft
drilled into the Pedro Miguel and Cucaracha formations, February 2003.
0
5
0 5,000 10,000 15,000 20,000 25,000 30,000
LOAD (KN) ß
0.30
0.45
5858
10
15
20
25
30
DEPTH(m)
0.30
___
1.20
O-cell Tests on an 11 m
long, 460 mm square
precast concrete pile
driven in silica sand in
North-East Florida
(Data from McVay et al 1999)
0
2
4
6
8
0 500 1,000 1,500 2,000 2,500 3,000
Shaft Resistance, Rs (KN)
(m)
E-F
LCPC
Schmertmann
Dutch
Meyerhof
Beta
Tests
5959
(Data from McVay et al. 1999)
A study of Toe and
Shaft Resistance
Response to
Loading
10
12
14
16
18
20
DEPTH
The foregoing analysis results are quite good predictions
They were performed after the test results were known
Such “predictions” are always the best!
So, what about true predictions?
6060
Let’s see the results of a couple of
Prediction Events
p

3/24/2013
11
U
L
T
I
M
A
T
E
R
Prediction Event at Deep Foundations
Institute Conference in Raleigh, 1988
6161
44 ft embedment,
12.5 inch square
precast concrete
driven through
compact silt and
into dense sand
Capacity in Static Loading Test = 200 tons
R
E
S
I
S
T
A
N
C
E
Tons
PREDICTORS (60 individuals)
1,500
2,000
2,500
ity(KN)
Orlando 2002 Predictions
Max Load
Available
6262
0
500
1,000
Predictors
Capac
500
600
700
KN)
0 20 40 60 80
MOVEMENT (mm)
FHWA Washington, DC, 1986
273 mm diam. closed-toe pipe pile driven 9.1 m into hydraulic sand fill
6363
0
100
200
300
400
1 2 3 4 5 6 7 8 9 10
PREDICTIONS
CAPACITY(
800
1,000
1,200
KN) 0 2 4 6 8 10 12 14 16 18
MOVEMENT (mm)
FHWA Baltimore, MD, 1980
Two 273 mm diam. closed-toe pipe piles driven 13.1 m into
Beaumont clay
6464
0
200
400
600
800
PARTICIPANTS
CAPACITY(K
1,500
2,000
2,500
3,000
3,500
OAD(KN)
Singapore 2002
1,400
1,600
1,800
2,000
65
0
500
1,000
1 3 5 7 9 11 13 15 17 19 21 23 25 27 29 31 33
L
400 mm H-Pile (168 kg/m) driven through
sandy clay to a 15 m embedment
0
200
400
600
800
1,000
1,200
0 10 20 30 40 50
MOVEMENT (mm)
LOAD(KN)
Brazil 2004: Bored pile (Omega screw pile) 23 m long, 310 mm diameter
0
2
4
6
8
0 20 40 60 80
Water Content (%)
(m)
0
5
10
0 5 10 15 20 25
N-Index (blows/0.3)
(m)
SPT 18
at 23 m Pile
0
2
4
6
8
0 25 50 75 100
Grain Size (%)
(m)
SILT
SAND CLAY
Sandy Silty
CLAY
(Laterite)
Sandy SILT
6666
10
12
14
16
18
20
DEPTH
wnwP wLGW
15
20
25
DEPTH
10
12
14
16
18
20
DEPTH
Sandy SILT
and CLAY
Sandy Clayey
SILT
GW

3/24/2013
12
Brazil 2004
Static Loading Test
on a 23 m 310 mm bored pile
Load-Movement Response
1,500
2,000
2,500
KN)
Prediction Compilation
2,000
2,500
PUSH L= 23m
0 5 10 15 20 25 30
MOVEMENT (mm)
6767
0
500
1,000
0 10 20 30 40
MOVEMENT (mm)
LOAD(K
0
500
1,000
1,500
PARTICIPANTS
LOAD(KN)
Portugal 2004. Precast 350 mm diameter pile driven to 6 m depth
in a saprolite, a residual soil consisting of silty clayey sand.
0
1
2
3
0 10 20
)
1 500
2,000
2,500
3,000
PACITY(KN)
CAPACITY FROM STATIC LOADING TEST
Pile C1
6868
4
5
6
7
8
DEPTH(m
0
500
1,000
1,500
1
PREDICTIONS
TOTALCAP
0
OFFSET LIMIT LOAD
1,200
1,400
1,600
1,800
KN)
Pipe-Pile
0 10 20 30 40
MOVEMENT (mm)
Northwestern University, Evanston, Illinois, 1989.
15 m embedment, 457 mm diameter closed-toe pipe piles driven in sand on clay.
6969
0
200
400
600
800
1,000
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24
CAPACITY(K
PREDICTIONS
Finno 1989
Edmonton, Alberta, 2011
Prediction of load-movement and capacity of a 400-mm diameter, 18 m
long, augercast pile constructed in transported and re-deposited glacial till.
2 000
3,000
4,000
D(KN)
E = 20 GPaE = 35 GPa
70
0
1,000
2,000
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT (mm)
LOAD
10 capacity predictions are
at movements > 50 mm
7 mm (4 mm + b/120 mm)
TEST
RESULTS
Values
RealtiveFrequency
2σ
σ = Standard
Deviation,
σ = 833
µ = Mean, 1,923
σ/µ = Coefficient
of Variation,
COV = 0.43
Mean, µ
4σ
NORMAL DISTRIBUTION
Edmonton 2011
71
CAPACITY PREDICTIONS
T he area between -σ and +σ fro m the mean value is 68% o f to tal area
T he area between -2σ and +2σ fro m the mean value is 95% o f to tal area
T he area between -3σ and +3σ fro m the mean value is 99% o f to tal area
0
100
200
0 10 20 30 40 50 60 70 80 90 100
PREDICTEDLOAD=100
MOVEMENT (mm)
NORMALIZED TO LOAD
Forthcoming Prediction Event in Bolivia April 2013
Four bored instrumented piles in sand tested in compression
0
5
2.9 m GW
1.0 m
4.5 m
1.0 m
4.5 m
Ground
surface
TP1 TP2 TP3 TP4
"Std" FDP FDP "Std"
+EB +O-cell
+EB
BH1 BH3 BH 4 BH2
0.0 m 400 mm 440 mm 400 mm 400 mm
72
5
10
15
20
25
DEPTH(m)
1.2 m17.5 m 2.5 m
15.0 m
O-cell
EB EB
600 mm
5
7.5 m
10.5 m
13.5 m
16.5 m 15.8 m
4.5 m
7.5 m
10.5 m
13.5 m
Test Pile Configurations and Strain-Gage Levels
440 mm400 mm 600 mm

3/24/2013
13
0
2
4
6
8
10
0 10 20 30 40 50
N (blows/0.3m)
PTH(m)
SPT1
SPT2
SPT3
0
2
4
6
8
10
0 5 10 15 20 25 30
WATER CONTENT (%)
TH(m)
0
2
4
6
8
10
0 20 40 60 80 100
GRAIN SIZE (%)
TH(m)
Fine to Medium Sand
Medium to
Coarse SandFines
BH-1
Soil Profile
73
12
14
16
18
20
DEP
10
12
14
16
18
20
DEPT
10
12
14
16
18
20
DEPT Gravel
Zone of Clay and Clayey Sand
(no samples)
Deadline for submitting a prediction is April 1
I will be glad to email the details for how to submit one.
Pore Pressure Dissipation
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
0
5
10
0 100 200 300 400 500 600
PORE PRESSURE (KPa)
(m)
7474
Paddle River, Alberta, Canada
(Fellenius 2008)
15
20
25
DEPTH
Before
Driving
EOID
Total
Stress
15
20
25
DEPTH
30 Days
after EOID 15 Days
after EOID
Before
Driving
EOID
Total
Stress
15
20
25
DEPTH
4 Years
after Driving
30 Days
after EOID 15 Days
after EOID
Before
Driving
EOID
Total
Stress
800
1,000
1,200
1,400
1,600
D(KN)
Effective Stress Analysis
0
5
10
0 500 1,000 1,500 2,000
LOAD (KN)
(m)
4 Years
after EOID
7575
0
200
400
600
0 10 20 30 40 50
MOVEMENT (mm)
LOA
Paddle River, Alberta, Canada
15
20
25
DEPTH(
15 Days
after EOID
30 Days
after EOID
All three analyses apply the same
coefficients coupled with the actual
pore pressure distribution
If we want to know the load distribution, we
can measure it. But, what we measure is
the increase of load in the pile due to the
load applied to the pile head. What about
the load in the pile that was there before
76
p
we started the test?
That is, the Residual load.
Normalized Applied Load
Load distributions in
static loading tests on
four instrumented
77
D
E
P
T
H
piles in clay
S d
Example from Gregersen et al., 1973
0
2
4
6
8
0 50 100 150 200 250 300
LOAD (KN)
(m)
0
2
4
6
8
0 100 200 300 400 500 600
LOAD (KN)
(m)
True
Residual
True minus
Residual
78
B. Load and resistance in DA
for the ultimate load applied
Sand8
10
12
14
16
18
DEPTH(
Pile DA
Pile BC,
Tapered
8
10
12
14
16
18
DEPTH(
A. Distribution of residual load in DA and BC
before start of the loading test

3/24/2013
14
FHWA tests on 0.9 m diameter bored piles
One in sand and one in clay
(Baker et al., 1990 and Briaud et al., 2000)
0
2
4
0 10 20 30 40
Cone Stress and SPT N-Index
(MPa and bl/0.3 m)
Silty
Sand
0
2
4
0 10 20 30 40
Cone Stress (MPa)
ClaySilty
Sand Clay
79
6
8
10
12
DEPTH(m)
Sand
Pile 4
6
8
10
12
DEPTH(m)
Pile 7
N
qc
Sand Clay
ANALYSIS RESULTS: Load-transfer curves
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
)
True
Distribution
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
Measured
Distribution
0.0
2.0
4.0
0 1,000 2,000 3,000 4,000 5,000
LOAD (KN)
m)
True
Distribution
Residual
Load
80
6.0
8.0
10.0
12.0
DEPTH(m
PILE 4
SAND
Measured
Distribution6.0
8.0
10.0
12.0
DEPTH(m)
PILE 4
SAND
Residual
Load
Measured
Distribution
6.0
8.0
10.0
12.0
DEPTH(m
PILE 7
CLAY
6.0
8.0
10.0
12.0
DEPTH(m
PILE 7
CLAY
Results of analysis of a Monotube pile in sand
(Fellenius et al., 2000)
0
5
0 1,000 2,000 3,000
LOAD (KN)
Measured
Resistance
Residual
Load
81
10
15
20
25
DEPTH(m)
True
Resistance
Method for evaluating
the residual load distribution
0
2
4
0 500 1,000 1,500 2,000
RESISTANCE (KN)
Measured
Load
Shaft
82
6
8
10
12
14
16
DEPTH(m)
Measured Shaft
Resistance
Divided by 2
Residual
Load
True
Resistance
Extrapolated
True Resistance
Resistance
0
5
10
15
20
0 500 1,000 1,500 2,000 2,500
LOAD (KN)
(m)
Static Loading Test
at Pend Oreille, Sandpoint, Idaho, for
the realignment of US95
406 m diameter,
45 m long, closed-toe pipe pile
driven in soft clay
Determining True Resistance
from Measured Resistance (“False Resistance”)
Cl
83
25
30
35
40
45
50
DEPTH(
Fellenius et al.
(2004)
driven in soft clay
200+ m
Clay
0
5
10
15
20
-500 0 500 1,000 1,500 2,000
LOAD (KN)
(m)
ß = 0.60
ß = 0.06
AS MEASURED,
i.e. "FALSE RES."
A
ß = 0.09
0
5
10
15
20
-500 0 500 1,000 1,500 2,000
LOAD (KN)
(m)
ß = 0.60
ß = 0.09
ß = 0.09
AS MEASURED,
i.e. "FALSE RES."
CPTu
Eslami-Fellenius
B
84
Test on a strain-gage instrumented, 406 mm diameter,
45 m long pile driven in soft clay in Sandpoint, Idaho
25
30
35
40
45
50
DEPTH
ß = 0.06
"TRUE RES."RESIDUAL
LOAD
AFTER 1st
UNLOADING
25
30
35
40
45
50
DEPTH
ß = 0.10
"TRUE RES."
per CPTu
RESIDUAL
LOAD
AFTER 1st
UNLOADING
ß = 0.10
Extrapolated

3/24/2013
15
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
PTH(m)
True
Resistance
HEAD-DOWN
AND FULL
RESIDUAL LOAD
Residual
Load
True
Resistance
False
Resistance
Silty
Sand
Silty
Clay
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
PTH(m)
HEAD-DOWN
AND PARTIAL
RESIDUAL LOAD
True
False
Resistance
Shaft
Resistance
Typical Example: Table 7.3 in the Red Book
85
20
25
30
35
DEP
Resistance
Residual and True
Toe Resistance
Transition
Zone
Silty
Sand
Glacial
Till
20
25
30
35
DEP
Residual
Load
Resistance
Residual and True
Toe Resistance
Transition
Zone
Resistance
The effect of residual load on an uplift test
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
True
Resistance
TENSION TEST
AND FULL
RESIDUAL LOAD
Residual
Load
0
5
10
-2,000 -1,500 -1,000 -500 0 500 1,000
LOAD (KN)
m)
Residual
Load
True
Resistance
TENSION TEST
AND PARTIAL
RESIDUAL LOAD
8686
15
20
25
30
35
DEPTH(m
False
Resistance
Toe Resistance
in an Uplift Test?!
15
20
25
30
35
DEPTH(m
False
Resistance
Toe Resistance
in an Uplift Test?
Combining the results of a head-down test with those of a
tensions test will help determining the true resistance
0
5
10
15
0 500 1,000 1,500 2,000 2,500 3,000 3,500
LOAD (KN)
H(m)
HEAD-DOWN
AND PARTIAL
RESIDUAL LOAD
False
Head-down
True Shaft
False
Tension
Test
8787
20
25
30
35
DEPTH
Residual
Load
True
Resistance
Residual and True
Toe Resistance
Transition
Zone
True Shaft
Resistance
Not directly
useful below
this level
Now you know why some claim that resistance in
tension is smaller than that in compression
400
600
800
1,000
LOAD(KN)
No Residual
Load
Residual Load
present
No Strain Softening
Presence of residual load is not just of academic interest
400
600
800
1,000
LOAD(KN)
With Strain Softening Residual
Load present
No Residual
Load
8888
0
200
400
0 5 10 15 20 25 30
MOVEMENT (mm)
L
OFFSET
LIMIT LOAD
0
200
400
0 5 10 15 20 25 30
MOVEMENT (mm)
L
OFFSET
LIMIT LOAD
• "Residual Load " follows the same principle and
mechanism as "Drag Load". The distinction made
is that by residual load we mean the locked-in load
present in the pile immediately before we start a
static loading test. By drag load we mean the load
present in the pile in the long-term.
Additional Comments on Residual load
8989
• Residual load as well as drag load can develop in
coarse-grained soil just as it does in clay soil.
• Both residual load and drag load develop at
very small movements between the pile and the
soil.
600
800
1,000
1,200
D(KN)
HEAD
TOE TELLTALE
A
Does not this shape of
Residual Load Affects Toe Resistance
Response
9090
0
200
400
600
0 5 10 15 20 25
MOVEMENT (mm)
LOAD
TOE
Does not this shape of
measured toe movement
suggest that there is a
distinct toe capacity?

3/24/2013
16
400
600
800
1,000
1,200
LOAD(KN)
HEAD
TOE
TOE TELLTALE
A
400
600
800
1,000
1,200
LOAD(KN)
HEAD
TOE
B
9191
0
200
0 5 10 15 20 25
MOVEMENT (mm)
0
200
0 5 10 15 20 25
MOVEMENT (mm)
"Virgin" Toe Curve
No, it only appears that way when we forget to consider the residual
toe load (also called the initial, or “virgin” toe movement)
Miscellaneous Details
Open vs. Closed Toe
Tapered section
H section
9292
H-section
. . . . . . .
Special Conditions
Step-tapered pile
9393
"Add-on" toe
resistance
acting on a
donut-
shaped
area
Special Conditions
Step-tapered pile
Smooth-tapered pile
Conical pile (wood pile)
Calculate in
elements
(increments) at
t
9494
"Add-on" toe
resistance
acting on a
donut-
shaped
area
every metre or so
the shaft resistance
acting along the pile
and toe resistance
for the “donut” of
each element
Just because the design assumes that the pile shaft is
smooth and straight with parallel sides does not mean it is.
9595
A A
B B
A-A and B-B
The "donut" area A
minus B projection
acting like an extra
Pile Toe
An unintentional effect
for many bored piles
and intentional for
“multi-underreamed”
piles
9696

3/24/2013
17
9797 9898
PILES FOR AN EXPANSION OF A
LOADING DOCK
9999
CALCULATION OF PILE CAPACITY
and
LOAD-TRANSFER CURVES
355 mm diameter closed-toe pipe pile to 32 m embedment
Area, As = 1.115 m2/m Live Load, Ql = 200 KN
Area, At = 0.099 m2 Dead Load, Qd = 800 KN
SILT
CLAY
4 m
W
5 m
100100
, t , d
LAYER 1 Sandy Silt ρ = 2,000 kg/m β = 0.40
LAYER 2 Soft Clay ρ = 1,700 kg/m3 β = 0.30
LAYER 3 Silty sand ρ = 2,100 kg/m3 β = 0.50
With artesian head of 5 m
LAYER 4 Ablation Till ρ = 2,200 kg/m3 β = 0.55
Nt = 50
TILL
SAND
27 m
21 m
32 m
CALCULATION OF LOAD TRANSFER
Area, As = 1.115 m2/m Live Load, Ql = 200 KN Shaft Resistance, Rs = 1,817 KN
Area, At = 0.099 m2 Dead Load, Qd = 800 KN Toe Resistance, Rt = 1,205 KN
Total Load, Qa = 1,000 KN Total Resistance, Ru = 3,021 KN
F.S. = 3.02 Depth to N. P. = 26.51 m Load at N. P., Qmax = 1,911 KN
DEPTH TOTAL PORE EFFECTIVE INCR. Qd+Qn Qu-Rs
STRESS PRES. STRESS Rs
(m) (KPa) (KPa) (KPa) (KN) (KN) (KN)
LAYER 1 Sandy Silt ρ = 2,000 kg/m3 β = 0.40
0.00 30.00 0.00 30.00 0.0 800 3,021
1.00(GWT) 48.40 0.00 48.40 17.5 817 3,004
4.00 104.30 30.00 74.30 82.1 900 2,922
LAYER 2 Soft Clay ρ = 1,700 kg/m3 β = 0.30
4.00 104.30 30.00 74.30 900 2,922
6.00 136.04 57.06 78.98 26.0 951 2,870
8.00 168.08 84.12 83.96 27.7 1,005 2,816
10 00 200 37 111 18 89 20 29 4 1 063 2 758
101101
10.00 200.37 111.18 89.20 29.4 1,063 2,758
12.00 232.88 138.24 94.64 31.2 1,125 2,697
14.00 265.55 165.29 100.26 33.1 1,190 2,631
16.00 298.38 192.35 106.03 35.0 1,259 2,562
18.00 331.33 219.41 111.92 37.0 1,332 2,489
20.00 364.40 246.47 117.93 39.0 1,409 2,413
21.00 380.97 260.00 120.97 40.0 1,449 2,373
LAYER 3 Silty sand = 2,100 kg/m3 β = 0.50
21.00 380.97 260.00 120.97 1,449 2,373
23.00 422.17 280.00 142.17 76.3 1,596 2,226
25.00 463.45 300.00 163.45 88.2 1,766 2,055
27.00 504.80 320.00 184.80 100.1 1,960 1,861
LAYER 4 Ablation Till = 2,200 kg/m3 β = 0.55
27.00 504.80 320.00 184.80 1,960 1,861
30.00 569.93 350.00 219.93 372.4 2,332 1,489
32.00 613.41 370.00 243.41 285.1 2,617 1,205 for Nt = 50
0
5
10
0 1,000 2,000 3,000 4,000
LOAD and RESISTANCE (KN)
Qd + qn
Qd
Qallow
Qlive Qu
Plot of the Calculated Values
zs cr '' βσ+=
Dztt Nr == 'σ
Calculation of shaft and
toe resistance per the
effective stress method
102102
15
20
25
30
35
DEPTH(m)
Qu - rs
Rt
dzcAdzrAR zssss )''( βσ+∫=∫=
Dzttttt NArAR === 'σ
Mother Nature no like no
kinkie stuff

3/24/2013
18
0
5
10
15
0 1,000 2,000 3,000 4,000
LOAD and RESISTANCE (KN)
H(m)
103103
20
25
30
35
DEPTH
Transition
Zone
Qn
Note, just because we carried the static loading test to a
certain toe movement does not mean that Nature will
impose the same toe load and toe movement for the
long-term condition.
0
5
10
0 500 1000 1500 2000
LOAD
Qult/ RultQdead
0
5
10
0 500 1000 1500 2000
LOAD
Qult/ RultQdead
104104
A) Small settlement only in the surrounding soils B) Large settlement in the surrounding soils
15
20
25
DEPTH
Rs
Qn
(Rt)
15
20
25
DEPTH
(Rt) Rs
Qn
RESIDUAL LOAD
0
5
0 500 1000 1500 2000
LOAD
Qult/ Rult
A test pile.
Before the start of
the test there is no
105105
10
15
20
25
DEPTH
Residual
Toe Load
load on the pile head
A Case history of evaluation of static and dynamic tests on a
300 mm, 12 m long pile driven in sand. Data from Axelsson (2000).
GW
Silty CLAY
SAND with lenses of
clay and silty clay
Uniform SAND
(80% sand size)
with occasional
9.25"
235 mm
0m
2.5m
T E S T S
Static loading test 5 days after driving at Depth 12.8 m
Restrike after static test to final depth 13.0 m with PDA/CAPWAP
106106
with occasional
lens of Silty CLAY
13.0
Redrive to 13.0 m depth
Static loading test 1 day after redrive
Static loading test 8 days after redrive
Static loading test 120 days (4 months) after redrive
Static loading test 670 days (22 months) after redrive
Total unit weight 0 m - 2.5 m = 18 KN/m
3
Total unit weight 2.5 m - 13.0 m = 19 KN/m
3
Hydrostatic pore pressure distribution
cnt.
107107
0
1
2
3
0.00 0.20 0.40 0.60 0.80 1.00
Equivalent ß (- - -)
0
1
2
3
0 25 50 75 100
Unit Shaft Resistance (KPa)
0
1
2
3
0 25 50 75 100
Unit Shaft Resistance (KPa)
ß-Method E-F Method
Equivalent ß-coefficient
from CPTU sounding and
Eslami-Fellenius Method
Unit Shaft Resistance from
Equivalent ß-coefficient and CPTU
Method plus LCPC-Method
cnt.
108108
4
5
6
7
8
9
10
11
12
13
DEPTH(m)
E-F Method
4
5
6
7
8
9
10
11
12
13
DEPTH(m)
4
5
6
7
8
9
10
11
12
13
DEPTH(m)
LCPC Method

3/24/2013
19
250
300
350
400
450
500
(KN)
Static test 8 days
after driving
PDA/CAPWAP
after static test
Load-movement curves from a static loading test and
the CAPWAP-determined load-movement curve from a
subsequent same-day dynamic test.
cnt.
109109
0
50
100
150
200
250
0 5 10 15 20 25 30 35 40 45 50
MOVEMENT (mm)
LOAD Load-Movement Curves for static tests after the redrive
cnt.
250
300
350
400
450
500
OAD(KN)
1 Day
8 Days
4 Months
4 Months
(Reloading)
22 Months
110
0
50
100
150
200
0 10 20 30 40 50 60 70
MOVEMENT (mm)
LO
An obvious example of set-up in sand — Right?
300
350
400
450
500
KN)
1 Day
8 Days
cnt.
When plotting the data in sequence as the tests
progressed from unloadings to reloadings, no
time-dependent increase can be discerned.
111
0
50
100
150
200
250
0 25 50 75 100 125 150 175 200
MOVEMENT (mm)
LOAD(K
8 Days
4 Months
4 Months
(Reloading)
22 Months
100
150
200
OELOAD(KN)
1 Day
8 Days
4 Months
22 Months
Toe load from earth stress cell at pile toe
cnt.
112
0
50
0 50 100 150
MOVEMENT OF PILE HEAD (mm)
TO
This indicates an ultimate toe resistance, i.e., no increase
of toe resistance for increasing toe movement — Right?
Toe load from earth stress cell at pile toe
cnt.
100
150
200
INCREASE(KN)
1 Day
8 Days
4 Months
22 Months
113
Residual
Toe Load
The entire history of the toe response needs to be considered.
A plot of entire history does not show an ultimate value.
Residual load can be determined from instrumented tests.
0
50
0 25 50 75 100 125 150
MOVEMENT (mm)
LOAD
Redundancy is nothing to look down on
114

3/24/2013
1
BASICS OF DESIGN
OF PILED
FOUNDATIONS
B t H F ll iBengt H. Fellenius
The Static Loading Test
Performance, Instrumentation, Interpretation
Bolivia, April 25, 2013
33
Candidates for Darwin Award, First Class
44
! ! !
Testing piles is a
risky business.
55
! ! ! 2 SPACER
1. SWIVEL
PLATE
What do you
think could
happen to the
stack of four
pieces on the
pile head when
66
4. JACK
3. LOAD CELL
2. SPACER the load is
applied? And,
therefore, to the
three oblivious
persons next to
the pile?

3/24/2013
2
77This is how experience taught the three, and others, to arrange the units on the pile head 88
99 1010
1111 12

3/24/2013
3
1313 14
Fellenius 1984
250
300
350
d(KN)
Head-down
O-cell Pile
August 2006
The error can be small or it can be large. Here are results
from two tests at the same site using the same equipment
testing two adjacent piles, one after the other.
1,500
2,000
)
15% Error
Shinho-Pile August 2006
15
0
50
100
150
200
0 2,000 4,000 6,000 8,000 10,000
Loadcell (KN)
ErrorinJackLoad
2.5% Error
0
500
1,000
0 5,000 10,000 15,000
Loadcell load (KN)
Error(KN)
2.5% Error
Note, the test on the pile called "O-cell pile" is a head-down test after a preceding O-cell test.
A routine static loading test provides
the load-movement of the pile head...
and the pile capacity?
16
The Offset Limit Method
Davisson (1972)
L
L
EA
Q Δ=
Q
17
OFFSET (inches) = 0.15 + b/120
OFFSET (SI-units—mm) = 4 + b/120
b = pile diameter (inch or mm)
LΔ
The Decourt Extrapolation
Decourt (1999)
1,000,000
1,500,000
2,000,000
--Q/s(inch/kips)
1
2
C
C
Qu =
C1 = Slope
C2 = Y-intercept
δ
Q
18
0 100 200 300 400 500
0
500,000
, ,
LOAD (kips)
LOAD/MVMNT-
Ult.Res = 474 kips
Linear Regression Line
Q

3/24/2013
4
Other methods are:
The Load at Maximum Curvature
Mazurkiewicz Extrapolation
Chin-Kondner Extrapolation
19
DeBeer double-log intersection
Fuller-Hoy Curve Slope
The Creep Method
Yield limit in a cyclic test
For details, see Fellenius (1975, 1980)
DECOURT 235
20
1,500
2,000
2,500
N)
Definition of capacity (ultimate resistance) is only needed
when the actual value is not obvious from the load-
movement curve
21
0
500
1,000
0 5 10 15 20 25 30 35 40
MOVEMENT (mm)
LOAD(KN
Offset-
Limit
Line
The capacity is not a constant, but changes with time
1,500
2,000
2,500
3,000ACITY(KN) 8 years
BOR
16 h BOR
48 days
Static
Test
CASE 1
4 years20 m
22
0
500
1,000
0.01 0.10 1.00 10.00 100.00 1,000.00 10,000.00
DAYS AFTER EOID
CAPA
1 h BOR
EOIDs CASE 2
16 m
Pile Toe Movement
3,000
4,000
AD(KN)
HEAD
HEAD LOAD vs. TOE MOVEMENT
65 ft long, 14 inch
pipe pile.
With a telltale to the
toe arranged to
23
0 5 10 15 20 25 30
0
1,000
2,000
MOVEMENT (mm)
LOADATPILEHEA
determine pile
shortening. Don’t
arrange it to
measure toe
movement directly.
Analysis of toe resistance
An adjacent pull test on a
similar pile established
that the pile shaft
resistance (2,000 KN)
was approximately fully
3,000
4,000
D(KN)
HEAD
HEAD LOAD vs. TOE MOVEMENT
PILE SHORTENING
24
was approximately fully
mobilized just short of a
5-mm upward movement
at the pile toe. Therefore
the load applied in the
push test beyond a toe
movement of 5 mm goes
to toe resistance, only.0 5 10 15 20 25 30
0
1,000
2,000
MOVEMENT (mm)
LOADATPILEHEAD
ESTIMATED TOE LOAD
vs.
TOE MOVEMENT
(Based on the assumption that
shaft resistance is 2,000 KN)
+10 %
-10 %

3/24/2013
5
20 inch square diameter, prestressed concrete pile driven to
58 ft embedment, through about 45 ft of soft silt and clay, 5
ft of sand, and to bearing 6 ft into hard clay
PUSH
and
PULL
To separate
f
Unloading-
reloading
once or a
couple of
times “on
the way up”
400
500
600
)
Push test
Offset LimitTOE
HEAD
25
Data from AATech Scientific Inc.
shaft and toe
resistances.
The pile is
equipped with
a toe telltale.
y p
serves no
purpose and
may result
in distorted
analysis
results
0
100
200
300
0.0 0.2 0.4 0.6 0.8 1.0 1.2
MOVEMENT (in)
LOAD(kips
Pull test
Combining the push and pull test results with the telltale
measurements to determine the load-movement for the pile toe
400
500
600
ps)
PUSH
TEST
TOE
"Toe
Telltale "
26
Data from AATech Scientific Inc.
0
100
200
300
0.0 0.2 0.4 0.6 0.8 1.0 1.2
MOVEMENT (in)
LOAD(ki
PULL
TESTSHAFT
From pull test with the
head movement
adjusted to the toe
movement
Instrumentation
a d
27
and
Interpretation
T e l l t a l e s
• A telltale measures shortening of a pile and must never be arranged to
measure movement.
• Let toe movement be the pile head movement minus the pile shortening.
• For a single telltale, the shortening divided by the distance between
the pile head and the telltale toe is the average strain over that length.
• For two telltales, the distance to use is that between the telltale tips.
28
, p
• The strain times the cross section area of the pile times the pile material
E-modulus is the average load in the pile.
• To plot a load distribution, where should the load value be plotted?
Midway of the length or above or below?
Load distribution for constant unit shaft resistance
0
0 100LOAD, Q
A1
Average
Load0
PILE HEAD
29
50
100h
A2
Midheight
Load
Distribution
Q = az
DEPTH, z
PILE TOE
ars =
21 AA =
Linearly increasing unit shaft resistance
and its load distribution
0
0 Unit Shaft
Resistance
az 3
1
3
ax
A =
)23( 323
xhxha +
0
0 Average
Load
LOAD, Q0
A1
x
30
h
DEPTH, z
21 AA =
h
h
X 58.0
3
==“X” is where the average load should be plotted
6
)23(
2
xhxha
A
+−
=
A2
Load
Distribution
Q = az2
/2
h

3/24/2013
6
• Today, telltales are not used for determining strain (load) in a pile
because using strain gages is a more assured, more accurate, and
cheaper means of instrumentation.
• However, it is good policy to include a toe-telltale to measure toe
movement. If arranged to measure shortening of the pile, it can also be
used as an approximate back-up for the average load in the pile.
Th f ib ti i t i ( ti l t i l
31
• The use of vibrating-wire strain gages (sometimes, electrical
resistance gages) is a well-established, accurate, and reliable means
for determining loads imposed in the test pile.
• It is very unwise to cut corners by field-attaching single strain gages to
the re-bar cage. Always install factory assembled “sister bar” gages.
32
Rebar Strain Meter — “Sister Bar”
Instrument Cable
Three
bars?!
Reinforcing Rebar
or Strand
Instrument Cables
33
Reinforcing Rebar
Rebar Strain Meter
Wire Tie
or Strand
Tied to Reinforcing Rebar
Hayes 2002
Wire Tie
Tied to Reinforcing Rings
(2 places)
Rebar Strain Meter
(3 places, 120° apart)
8
10
12
14
16
18
20
LOAD(MN)
Load-strain of individual gages and of averages
4
6
8
10
12
14
16
18
LOAD(MN)
LEVEL 1 D CA B
A&C B&D
34
0
2
4
6
0 50 100 150 200
STRAIN (µε)
Level 1A+1C
Level 1B+1D
Level 1 avg 0
2
4
0 100 200 300 400 500 600 700
STRAIN (µε)
The curves are well together and
no bending is discernable
Both pair of curves indicate bending; averages are very close;
essentially the same for the two pairs
If one gage “dies”, the data of surviving single gage should be discarded.
It must not be combined with the data of another intact pair.
Data from two surviving single gages must not be combined.
12
14
16
18
20
N)
A&C+D
Means: A&C, B&D, AND A&B&C&D
A&C+B
B+CA+D
LEVEL 1
35
0
2
4
6
8
10
0 100 200 300 400 500 600 700
STRAIN (µε)
LOAD(MN
Error when including
the single third gage, when
either Gage B or Gage D data
are discarded due to damage.
Glostrext Retrievable Extensometer (Geokon 1300 & A9)
36
Lee Sieng Kai, 2010. Recent development in pile instrumentation
technology for driven, jacked-in and bored cast-in-place piles.
Lecture notes. [www.glostrext.com.my]
Anchor arrangement display Anchors installed

3/24/2013
7
Gage for measuring
displacement, i.e., distance
change between upper and
lower extensometers.
Accuracy is about 0.02mm/5m
37
corresponding to about 5 µε.
That the shape of a pile sometimes can be quite different from the straight-sided cylinder can be noticed in a
retaining wall built as a pile-in-pile wall
38
0
5
10
0.00 0.50 1.00 1.50 2.00 2.50
DIAMETER RATIO AND AREA RATIO
(m)
Nominal
Ratio
Determining actual shape of the bored hole before concreting
39
15
20
25
DEPTH
Gage
Depth
Diameter
Ratio
Area
Ratio
O-cell

3/24/2013
1
We have got the strain.
How do we get the load?
• Load is stress times area
1
• Stress is Modulus (E) times strain
• The modulus is the key
εσ E=
For a concrete pile or a concrete-filled bored pile, the
modulus to use is the combined modulus of concrete,
reinforcement, and steel casing
cs
ccss
comb
AA
AEAE
E
+
+
=
2
Ecomb = combined modulus
Es = modulus for steel
As = area of steel
Ec = modulus for concrete
Ac = area of concrete
• The modulus of steel is 200 GPa (207 GPa for those weak at heart)
• The modulus of concrete is. . . . ?
Hard to answer. There is a sort of relation to the cylinder strength and the
modulus usually appears as a value around 30 GPa, or perhaps 20 GPa or
so, perhaps more.
This is not good enough answer but being vague is not necessary.
The modulus can be determined from the strain measurements.
3
Calculate first the change of strain for a change of load and plot the
values against the strain.
Values are known
ε
σ
Δ
Δ
=tE
50
60
70
80
90
100
DULUS(GPa)
Level 1
Level 2
Level 3
Level 4
Level 5
Example of “Tangent Modulus Plot”
4
0 200 400 600 800
0
10
20
30
40
50
MICROSTRAIN
TANGENTMO
Best Fit Line
ba
d
d
Et +=⎟
⎠
⎞
⎜
⎝
⎛
= ε
ε
σ
εεσ b
a
+⎟
⎞
⎜
⎛
= 2
Which can be integrated to:
B t stress is also a f nction of
In the stress range of the static loading test, modulus of concrete is
not constant, but a more or less linear relation to the strain
5
εεσ b+⎟
⎠
⎜
⎝
=
2
εσ sE=
But stress is also a function of
secant modulus and strain:
Combined, we get a useful relation:
baEs += ε5.0 and Q = A Es ε
50
60
70
80
90
100
DULUS(GPa)
Level 1
Level 2
Level 3
Level 4
Level 5
Example of “Tangent Modulus Plot”
6
0 200 400 600 800
0
10
20
30
40
50
MICROSTRAIN
TANGENTMO
Best Fit Line
Intercept is
”b”
Slope is “a”

3/24/2013
2
Note, just because a strain-gage has registered some strain
values during a test does not guarantee that the data are useful.
Strains unrelated to force can develop due to variations in the pile
material and temperature and amount to as much as about 50±
microstrain. Therefore, the test must be designed to achieve
strains due to imposed force of ideally about 500 microstrain and
7
beyond. If the imposed strains are smaller, the relative errors and
imprecision will be large, and interpretation of the test data
becomes uncertain, causing the investment in instrumentation to
be less than meaningful. The test should engage the pile material
up to at least half the strength. Preferably, aim for reaching close
to the strength.
Unlike steel, the modulus of concrete varies and depends on curing, proportioning,
mineral, etc. and it is strain dependent. However, the cross sectional area of steel in an
instrumented steel pile is sometimes not that well known.
y = -0.0013x + 46.791
45
50
55
60
STIFFNESS,EA(GN)
45
50
55
60
STIFFNESS,EA(GN)
EAsecant (GN) = 46.5 from tangent stiffness
EAsecant (GN) = 46.8 - 0.001µε from secant stiffness
8
30
35
40
0 100 200 300 400 500 600
STRAIN, με
SECANTS
30
35
40
0 100 200 300 400 500 600
STRAIN, με
TANGENT
y = 0.000x + 46.451
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = Q/ε
(Data from Bradshaw et al. 2012)
Pile stiffness for a 1.83 m diameter steel pile: open-toe pipe pile.
Strain-gage pair placed 1.8 m below the pile head.
Field Testing and Foundation Report, Interstate H-1, Keehi
Interchange, Hawaii, Project I-H1-1(85), PBHA 1979.
4
5
Q/∆ε(GN)
TANGENT STIFFNESS, ∆Q/∆ε
4
5
ε(GN)
SECANT STIFFNESS, Q/ε
Strain-gage instrumented, 16.5-inch octagonal prestressed
concrete pile driven to 60 m depth through coral clay and
sand. Modulus relations as obtained from uppermost gage
(1.5 m below head, i.e., 3.6b).
9
Data from PBHA 1979
y = -0.0014x + 4.082
0
1
2
3
0 500 1,000 1,500 2,000
STRAIN (µε)
TANGENTSTIFFNESS,∆Q
y = -0.0007x + 4.0553
0
1
2
3
0 500 1,000 1,500 2,000
STRAIN (µε)
SECANTSTIFFNESS,Q/ε
Secant Data
Secant from Tangent Data
Trend Line
10
15
TIFFNESS,EA(GN)
10
15
STIFFNESS,EA(GN)
y = -0.003x + 7.41
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = Q/ε
For the "calibrating" uppermost gage level, the secant
method appears to be the better one to use, right?
10
Pile stiffness for a 600 mm diameter concreted pipe
pile. The gage level was 1.6 m (3.2b) below pile head
Data from Fellenius et al. 2003
0
5
0 50 100 150 200 250 300
STRAIN, µε
SECANTST
0
5
0 50 100 150 200 250 300
STRAIN, µε
TANGENTS
y = -0.004x + 7.21
EAsecant (GN) = 7.2 - 0.002µε from tangent stiffness
y = -0.0053x + 11.231
20
30
40
50
NTSTIFFNESS,EA(GN)
20
30
40
50
ENTSTIFFNESS,EA(GN)
TANGENT STIFFNESS, EA = ∆Q/∆εSECANT STIFFNESS, EA = ∆Q/∆ε
Or this case? Here, that initial "hyperbolic" trend can
be removed by adding a mere 20 µε to the strain data,
"correcting the zero" reading, it seems.
11
0
10
0 100 200 300 400 500
STRAIN (µε)
SECAN
y = -0.0055x + 9.995
0
10
0 100 200 300 400 500
STRAIN (µε)
TANGE
Secant stiffness after adding
20µε to each strain value
Secant stiffness from
tangent stiffness
Pile stiffness for a 600-mm diameter prestressed pile.
The gage level was 1.5 m (2.5b) below pile the head.
Data from CH2M Hill 1995
Or the adding of a mere
8 µε for this case?
40
50
S,EA(GN)
40
50
SS,EA(GN)
Secant for
Virgin Loading
Trend Line from
Tangent Stiffness
0
2,000
4,000
6,000
8,000
10,000
12,000
14,000
0 1 2 3 4 5 6
MOVEMENT (mm)
LOAD(KN)
12
y = -0.008x + 30.295
10
20
30
0 100 200 300 400 500
STRAIN (με)
SECANTSTIFFNES
y = -0.0115x + 29.234
10
20
30
0 100 200 300 400 500
STRAIN (με)
TANGENTSTIFFNES
g
Relation
Secant Stiffness
after adding 8µε to
each strain value
Pile stiffness for a 900-mm bored pile constructed in Indonesia.
The gage level was 2.0 m (2.2b) below pile the head.

3/24/2013
3
After completion of the test, the pile
was reloaded. Below, the 2nd cycle data
have been added to the first cycle plot.
0
2,000
4,000
6,000
8,000
10,000
12,000
14,000
0 1 2 3 4 5 6
MOVEMENT (mm)
LOAD(KN)
40
50
SS,EA(GN)
40
50
ESS,EA(GN)
Secant for Reloading
(1st cycle strains removed)
13
Data from Geo Optima Pt. 2011
10
20
30
0 100 200 300 400 500
STRAIN (με)
SECANTSTIFFNES
10
20
30
0 100 200 300 400 500
STRAIN (με)
TANGENTSTIFFNE
Secant for Reloading
Tangent for Reloading
Illustration of the adverse effect of unloading/reloading.
What really do we learn from
unloading/reloading and what
14
unloading/reloading and what
does unloading/reloading do to
the gage records?
The Testing Schedule
150
200
250
300
ERCENT"
A much superior test schedule. It presents a large number of values (≈20 increments), has no
destructive unloading/reload cycles, and has constant load-hold duration. Such tests can be used
in analysis for load distribution and settlement and will provide value to a project, as opposed to the
long-duration, unloading/reloading, variable load-hold duration, which is a next to useless test.
Plan for 200 %, but make use of
the opportunity to go higher if
this becomes feasible
0
50
100
0 6 12 18 24 30 36 42 48 54 60 66 72
TIME (hours)
"PE
The schedule in blue is typical for many standards. However, it is costly, time-consuming,
and, most important, it is diminishes or eliminates reliable analysis of the test results.
XXXXX
What about keeping the load on the pile until "zero" movement?
(Long-duration load-holding)
30
40
50
60
D(MN)
Pile TP-1 Pile TP 1
30
40
50
60
(MN)
16
0
10
20
30
0 5 10 15 20 25 30
DAYS
LOAD
Lakhta Center, St Petersburg, Russia
2.0 m diameter, 84 m long, bored pile
0
10
20
30
0 20 40 60 80 100
MOVEMENT (mm)
LOAD
Pile TP 1
0
10
20
30
40
50
0 50 100 150 200 250 300
TIME (hours)
MOVEMENT(mm)
2L-13
2L-11
2L-10
2L-12
2L-14
Pile TP 1
0
5
10
15
20
25
30
35
0 5,000 10,000 15,000 20,000
TIME (minutes)
LOADBETEENGLs(MN)
GL6 to GL7 GL5 to GL6 GL5 to GL7
2L-14
2L-13
2L-12
2L-11
2L-14
2L-13
2L-12
2L-10
2L-122L-112L-10
2L-14
2L-10
2L-11
2L-13
17
Lakhta Center, St Petersburg, Russia
2.0 m diameter, 84 m long, bored pile
The long-duration load-holding and variations of load increments have obviously had
considerable costs consequence for the project. Yet, nothing was "bought" by those
costs. On the contrary, the uneven load-holding durations and the differing load increment
magnitudes messed up the data and reduced the usefulness of the detailed analysis of
the test records.
( )
TIME (minutes)
The occasional unloading/reloading and varying load-holding durations
provide no information of any value for assessing pile response to load. It
is nothing but a vestigial practice, i.e., remnant of old, now obsolete, part
of the practice, much like our tailbone.
Figuratively speaking, it is strange that so many still appear to believe that
they have a tail at their rear end to wag despite the fact that the vestigial
On unloading/reloading
18
they have a tail at their rear end to wag, despite the fact that the vestigial
tailbone is not connected to the head. Indeed, to schedule a test to
include unloading/reloading and varying load-holds duration is nothing but
akin to a insisting on that there is a tail to wag, disregarding all evidence
to the contrary. Those who argue for the wag seem to be too busy
contemplating their navel to realize that nothing useful happens.

Fellenius bases de diseño de pilotes de fundación

Fellenius bases de diseño de pilotes de fundación

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Fellenius bases de diseño de pilotes de fundación