ff

1. Cyclic and post-cyclic monotonic behavior of Adapazari soils
Zulkuf Kaya a,n
, Ayfer Erken b
a
Engineering Faculty, Department of Civil Engineering, Erciyes University, Turkey
b
Civil Engineering Faculty, Department of Civil Engineering, Istanbul Technical University, Turkey
a r t i c l e i n f o
Article history:
Received 23 December 2013
Received in revised form
25 March 2015
Accepted 3 May 2015
Available online 22 May 2015
Keywords:
Dynamic triaxial test
Cyclic behavior
Post-cyclic monotonic behavior
Adapazari soils
Kocaeli earthquake
a b s t r a c t
The August 17, 1999 Kocaeli earthquake affected the city of Adapazari, which is located in the northwest
of Turkey, with severe liquefaction and bearing capacity failures causing tilting of buildings, excessive
settlements and lateral displacements. To understand the stress–strain behavior and pore pressure
behavior of undisturbed soils during the earthquake, the cyclic and post-cyclic shear strength tests have
been conducted on soil samples obtained from Adapazari in a cyclic triaxial test system within the scope
of this research. Cyclic tests have been conducted under stress controlled and undrained conditions.
Post-cyclic monotonic tests have been conducted following cyclic tests. The strength curves obtained in
the experiments showed that the dynamic resistance of silty sand was found to be 45% lower than those
of high plasticity soils (MH). The strength of clayey soils with the plasticity index of PI¼15–16% was
lower compared to the strength of high plasticity soils. Also, it was observed that silty sand soils had the
lowest strength. The dynamic strength of the soils increased with the increase in plasticity.
& 2015 Elsevier Ltd. All rights reserved.
1. Introduction
The Kocaeli earthquake of Mw ¼7.4 magnitude occurred on
17 August 1999 in the northwestern part of Turkey at a depth of
about 17 km along the North Anatolian Fault zone. A right lateral
strike slip fault produced an extensive surface rupture over a
distance of 126 km, which was accompanied by the earthquake
[1]. The Kocaeli earthquake caused severe damage to hundreds of
structures and lifelines in Adapazari. A large number of modern
reinforced concrete buildings, generally 3–5 stories high, pene-
trated the surrounding ground or tilted due to liquefaction and
ground softening. Many of these buildings also had significant
structural damage [2]. Sand boils were observed within some of
the ground failure zones, but were not widespread and were
absent in many areas.
The effects of subsurface conditions on building damage and on
the occurrence of ground failure were investigated through
liquefaction assessment and one-dimensional site response ana-
lyses by various studies [3–10]. According to the results of these
studies, local variations in the characteristics of alluvial sediments
in Adapazari appear to have played a major role in the occurrence
of ground failures and been associated with building damages.
The liquefaction susceptibility and deformation behavior of
fine-grained soils were investigated through a series of standard
and rapid monotonic, and stress-controlled cyclic triaxial tests
conducted on undisturbed soil samples [11,12]. According to
Yılmaz et al. [11], the results of the series of tests showed that
the soils used did not display any trends that could be interpreted
as liquefaction regarding stiffness and strength response. Zehtab
[13] and Yılmaz and Zehtab [14] performed a series of cyclic direct
simple shear tests in order to assess dynamic properties of
Adapazari soils, which included different soil classes (CL, CH, ML,
SM). Also, the results of the tests conducted by Bray and Sancio
[12] showed that young, shallow, non-plastic silts and clayey silts
of low plasticity (PIo12) having ratios of water content to liquid
limit greater than 0.85 (wn/LL40.85) can liquefy under significant
cyclic loadings.
Assessment of stability, liquefaction and deformation behaviors
of such soils subjected to cyclic stresses requires the evaluation of
dynamic behavior. The purpose of this paper is to investigate and
evaluate the cyclic and post-cyclic behavior of Adapazari soils
(cohesive and cohesionless soils) under earthquake loading.
The cyclic laboratory tests conducted on undisturbed soil
samples obtained from Adapazari may enlighten the cyclic and
post-cyclic soil behavior in this scope. The results of a series of
tests designed to examine the stress–strain behavior and pore
pressure behavior of undisturbed soils under repeated loading are
presented. In this research dynamic tests will contribute to the
explanation of why buildings tilted, over-settled, or collapsed and
why lateral spreading of soil occurred as a result of the liquefaction
and ground softening during the Kocaeli earthquake in Adapazari.
For the determination of soil behavior by laboratory tests under
cyclic loading conditions, both disturbed and undisturbed soil
Contents lists available at ScienceDirect
journal homepage: www.elsevier.com/locate/soildyn
Soil Dynamics and Earthquake Engineering
http://dx.doi.org/10.1016/j.soildyn.2015.05.003
0267-7261/& 2015 Elsevier Ltd. All rights reserved.
n
Correspondence to: Engineering Faculty, Department of Civil Engineering,
Erciyes University, Kayseri 38039, Turkey.
E-mail addresses: zkaya@erciyes.edu.tr (Z. Kaya), erken@itu.edu.tr (A. Erken).
Soil Dynamics and Earthquake Engineering 77 (2015) 83–96

2. samples obtained from 10 borings at 8 different locations of
Adapazari were used. Two different sets of cyclic triaxial tests
were conducted on undisturbed soil samples. The first set included
tests where cyclic loading was continued until liquefaction
occurred (ε¼ 72.5% or ru ¼0.95–1.0), while the second set
involved the tests with different cyclic stress ratios applied for a
specific number of cycles, i.e. N¼20 was chosen to represent the
August 17, 1999 earthquake which had a moment magnitude of
Mw ¼7.4 at a frequency of 0.1 Hz. Strain controlled monotonic tests
with a loading speed of 0.20 mm/min were performed at the end
of cyclic loading application stage.
2. Geologic setting and local soil conditions
Most of the city is located over deep alluvial sediments. A deep
boring recently performed at the Yenigun District by the General
Directorate of State Water Works (DSI) did not reach bedrock at a
depth of 200 m. The shallow soils (or surface layers) are recent
deposits laid down by the Sakarya and Cark rivers, which fre-
quently flooded the area until flood control dams were built
recently. The sands accumulated along the bends of the mean-
dering rivers, and the rivers flooded periodically leaving behind
predominantly non-plastic silts, silty sands, and clays throughout
the city. Clay-rich sediments were deposited in lowland areas
where floodwaters ponded [15]. The thickness of the alluvial
deposits reaches around 1100 m below the center of the city of
Adapazari [6].
One of the reasons for the heavy damage to buildings, lifeline
systems, and bridges in this event was the widespread liquefaction
caused by the earthquake. The center of Adapazari suffered from
severe liquefaction and bearing capacity failure due to the softening
of fine grain soils. In this zone the observed settlements of many
buildings exceeded 1.0 m while settlements were limited on the
other side of the city located on the alluvial part. Furthermore, most
of the buildings have been tilted as a result of the liquefaction-
induced differential settlements. Besides liquefaction-related ground
deformation, deformations during bearing capacity failure also
developed in the center of the city.
Buildings in Adapazari were strongly shaken by the Kocaeli
earthquake. The Sakarya station recorded a peak horizontal (east–
west) ground acceleration (PGA), velocity, and displacement of
0.41g, 81 cm/s, and 220 cm, respectively. The Sakarya station is
located in southwestern Adapazari at a distance of 3.3 km from the
fault rupture. It is situated in a small one-storey building (with no
basement) and is underlain by a shallow deposit of stiff soil
overlying bedrock (average Vs¼470 m/s in the upper 30 m of the
soil) (Fig. 1). The central Adapazari is located at a distance of about
5–7 km from the fault rupture (Rrup) [2,13] and due to softer
ground conditions, amplification of long-period components of the
ground motion would be expected. The main shock ground
motions recorded at similar site-source distances suggest that
the PGA in Adapazari was on the order of 0.3–0.4 g [2].
3. Soil characterization in Adapazari
To investigate the subsurface conditions, twelve soil borings
with closely spaced SPT were drilled in Adapazari city where
ground failure was or was not observed. Undisturbed soil samples
were taken down to a depth of 20 m in every borehole, depending
on the soil thickness. The procedures outlined by ASTM standards
[16,17] were carefully followed to ensure the quality of the data.
BH 10 was drilled at the Sakarya Public Works and Settlement
Directorate Record Station; the other boreholes were drilled at
8 sites where buildings settled, tilted, or slid due to liquefaction or
ground softening (Fig. 2). The coordinates and the names of the
districts in which soil borings were performed and observed
damage during the earthquake at sites where more detailed
subsurface investigations conducted, are given in Table 1.
Tigcilar district, located downtown of the city, was the most
severely affected area from severe liquefaction and bearing capa-
city failure. Borehole BH4 is located on the foundation area of a
collapsed building in Tigcilar district as shown in Fig. 3. At this
location, a 4-story reinforced concrete building collapsed and the
surrounding buildings were tilted, settled and laterally moved
around 1.0 m. Corrected SPT-N values, (N1)60, were calculated as
suggested by Youd et al. [18]. Taking into consideration the average
borehole drilling experience in Turkey, the energy correction
coefficient (CE) was assumed to be 45/60¼0.75.
4. Test program
4.1. In situ tests and sampling
Subsurface soil conditions at the mentioned districts were
investigated through the analysis and interpretation of 12 explora-
tory borings with closely spaced SPT performed for this study.
4
8
12
16
20
0 10 20 30 40 50 60
SPT-(N)30 BH-10
Fill
Clay
Clay-Silt-Gravel
Sandstone
Schist
Schist
Fig. 1. Soil profile of borehole BH10 was drilled at the Sakarya Public Works and
Settlement Directorate Record Station.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
84

3. Organize
Hanlılar
Toyota
Serdivan
Tiğcılar
Cumhuriyet
Maltepe
Istiklal
Şeker
Semerciler
Yenidoğan
Yenigün
N
Istanbul
Ankara
Istanbul
Ankara
Scale :1/50 000
Main fault
Ozanlar
Tekeler
Sakarya
Yağcılar
Orta
Karaosman
Erenler
Güneşler
Sakarya Nehri
Papuçcular
:Locations of
Fig. 2. Map of the locations of SPT boring used in this study.
Table 1
Districts showing locations of SPT borings.
Borehole no. District Coordinate Explanation
N E
BH1 Yenidogan 40o
46.33 30o
23.55 PTT side
BH2 Papuccular 40o
46.24 30o
24.07 Turning building side (no liquefaction)
BH3 Yenigun 40o
46.38 30o
24.22 Bearing capacity loss
BH4 Tigcilar 40o
46.00 30o
24.00 Turning building side (liquefaction)
BH6 Cumhuriyet 40o
46.00 30o
23.70 in the face of Ataturk School (liquefaction)
BH7 Cumhuriyet 40o
46.00 30o
23.70 in the face of Ataturk School (liquefaction)
BH8 Semerciler 40o
46.00 30o
23.00 –
BH10 Record Station 40o
44.22 30o
23.04 Public Works and Settlement Directorate
BH11 Karaosman 40o
47.00 30o
23.00 –
BH12 Karaosman 40o
47.00 30o
23.00 –
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 85

4. 4.2. Laboratory tests
A series of laboratory tests, including index properties tests
(sieve analysis, specific gravity, liquid limit and plastic limit tests),
oedometer tests, cyclic and post-cyclic monotonic triaxial com-
pression tests were performed using the SPT samples and undis-
turbed Shelby tube samples to determine the properties and
strength characteristics of the soils obtained from boreholes
drilled in Adapazari city.
4.3. Soil properties
The physical properties of the specimens tested and cyclic
stress ratios (CSRs) are presented in Tables 2 and 3. The grain size
distributions of the samples tested for a group of SM, ML, and MH
are shown in Fig. 4; and those of the other group of CL and CH are
given in Fig. 5.
In this paper, liquefaction tests are referred to as “I. Group”,
cyclic loading tests and static tests following cyclic loading are
referred to as “II. Group”. In the I. Group tests, which are shown in
Table 2, the water content of the specimens is wn ¼29–50%, fines
content is FC¼39–98%, plasticity index is PI¼0–22%. The uncor-
rected SPT blow count of silty sand is 23 while the SPT value of soft
low plastic silt is 5. In the II. Group tests, samples have the values
of: wn ¼22–50%, FC¼1–100%, PI¼0–40% and wn ¼22–50%, and SPT
values are in the range of N30¼4–44.
Consistency index (Ic) is useful in the study of the field behavior
of saturated fine-grained soils (Tables 2–3). If the Ic of a soil is
equal to unity, it is at the plastic limit. Similarly, a soil with an Ic
equaling zero is at its liquid limit. The consistency index changes
from 0.333 to 0.762 for I. Group test samples. In the II. Group tests,
also, Ic changes from 0.500 to 1.000 except for the six soil samples
with Ic¼0.217–0.462. Generally, the consistency of the second
group test samples is expressed as medium-stiff and stiff.
4.4. Sample preparation
To minimize disturbance of fine-grained (clay, silty clay, silt
etc.) soils obtained from the districts of Adapazari city after Kocaeli
earthquake, Shelby tubes of approximately 7 cm in diameter and
80 cm in height were used. Sample preparation was performed in
accordance with the Japanese Geotechnical Standard [19].
Moreover, the soils (silt, silty sand, sandy silt, etc.), which had no
potential to stand in original form after being taken from the Shelby
tubes, were kept in the freezer along with the Shelby tubes as the
water was drained through holes made at the ends of the tubes.
Tubes were transversely divided into sections of about 12–15 cm.
Then, samples were retrieved from tubes which were split on both
4
8
12
16
20
24
28
32
0 10 20 30 40 50
SPT-(N1)60 BH-4
Fill
Clay
Silt
Silt
Clay
Silt
Silty Clayey Sand
S4
T3
Silty Sand
Silt
0 20 40 60 80
Consistency
wn
LL
PL
0 20 40 60 80 100
PI-FC (%)
PI
FC
Fig. 3. Soil profile and index properties of the borehole BH4.
Table 2
Physical properties of the samples used in the liquefaction experiments.
BH no. Test no. Depth (m) wn (%) FC (%) LL (%) PL (%) PI (%) Ic γs (g/cm³) N30 CSR Soil type
BH-3 S3-1 10.5–11.0 37 53.0 – – NP – 2.708 23 0.280 ML
BH-3 S3-2 10.5–11.0 29 46.0 – – NP – 2.632 23 0.205 SM
BH-3 S3-3 10.5–11.0 32 39.0 – – NP - 2.665 23 0.210 SM
BH-3 S3-4 3.0–3.5 46 92.0 62 41 21 0.762 2.690 5 0.331 MH
BH-3 S3-5 3.0–3.5 50 94.0 55 40 15 0.333 2.712 5 0.350 MH
BH-3 S3-6 3.0–3.5 46 94.0 59 39 20 0.650 2.730 5 0.377 MH
BH-11 T12-1 4.0–4.5 33 98.0 38 22 16 – – 11 0.346 CL
BH-12 T12-2 4.0–4.5 38 95.0 38 23 15 – – 12 0.315 CL
wn, water content; FC, fines content; LL, liquid limit; PL, plastic limit; PI, plasticity index; Ic, consistency index; γs, specific gravity; N30, SPT blow count value; and CSR, cyclic
stress ratio.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
86

5. sides equally by the planer in the longitudinal. The specimens were
prepared so that the length was double the width. Therefore, the
diameter was 50 mm, and the length was 100 mm. The prepared
test sample and testing apparatus are shown in Fig. 6.
According to the results of oedometer tests, the overconsolida-
tion ratios of soil samples taken from BH11 (2.0–2.5 m), BH11
(4.0–4.5 m) and BH12 (2.0–2.5 m) boreholes were determined as
3.85, 2.75, 3.70, respectively. The top soil layer to the depth of
4.5 m in Adapazari city is slightly overconsolidated. Such deter-
mined overconsolidation ratios are compatible with the results of
the experiments conducted by Bray and Sancio [12], and Sancio
et al. [20], and Sancio [21], and Erken and Ulker [22].
It is reported by Sancio [21] that “The actual stress history
profile in Adapazari somewhat uncertain; however, the shallow
soil deposits tested in his study are believed to have OCRs between
2 and 4, due primarily to desiccation, which has an important
effect on the stress state of the soil. Stresses in the pore water due
to desiccation that have a magnitude beyond the in-situ over-
burden stress, as is believed to be the case of Adapazari, will
induce hydrostatic (isotropic) consolidation. Therefore, the ratio
between the horizontal and vertical effective stresses is equal or
closes to one (which corresponds to isotropic consolidation) at this
stage.” Therefore, in this study the samples were loaded under
isotropic confining pressure.
5. Conducting of cyclic tests
The cyclic triaxial test is one of the most reliable and useful
laboratory tests for determining the stress–strain characteristics of
soils under dynamic loading. The test can simulate field conditions
and permits excess pore water pressure measurement so its
results can be accepted as more reliable than those of other tests
[23]. The cyclic triaxial compression tests were performed in
accordance with the Japanese Geotechnical Standard [24,25] with
the failure criterion defined as 5% axial strain.
The cyclic triaxial test systems located at the Istanbul Technical
University Soil Dynamic Laboratory were used in this study. Cyclic
triaxial strength test results are used for evaluating the ability of a
soil to resist the shear stresses induced in a soil mass due to
earthquake or other cyclic loading. The cyclic triaxial test is used
mostly to determine cyclic soil strength.
If soils (silt, silty sand, sandy silt, etc.) had the potential to stand
after being taken, the specimen was placed in the triaxial cell after
trimming. Then, the effective stress value computed according to
Table 3
Physical properties of samples in which strain controlled monotonic tests were performed at the end of 20 cyclic loading application stages.
BH no. Test no. Depth (m) wn (%) FC (%) LL (%) PL (%) PI (%) Ic γs (kN/m³) N30 CSR Soil type
BH4 S4-1 3.0–3.5 22 55.5 – – – – 26.34 12 0.320 ML
BH6 S6-1 3.0–3.5 45 97.3 50 27 23 0.217 27.70 7 0.420 CH
BH6 S6-2 3.0–3.5 42 98.8 59 22 37 0.459 27.50 7 0.482 MH
BH7 S7-1 3.0–3.5 41 98.1 74 41 33 1.000 26.96 4 0.370 MH
BH7 S7-2 3.0–3.5 44 97.0 79 43 36 0.972 27.78 4 0.420 MH
BH7 S7-3 3.0–3.5 43 99.3 66 35 31 0.742 27.40 4 0.515 MH
BH8 S8-1 3.0–3.5 37 86.6 – – NP – 25.94 5 0.240 ML
BH8 S8-2 3.0–3.5 41 89.5 – – NP – 26.25 5 0.255 ML
BH8 S8-3 3.0–3.5 44 97.3 – – NP – 27.15 5 0.400 ML
BH4 T3-1 6.0–6.5 22 97.5/9.2 40 16 24 0.750 – 15 0.320 CL/SP-SC
BH2 T4-1 4.5–5.0 – 1 – NP NP – – 44 0.408 SP
BH1 T5-1 5.0–5.5 38 98.9 66 26 40 0.700 – 5 0.400 CH
BH1 T5-2 5.0–5.5 32 98.4 40 24 16 0.500 – 5 0.340 CL
BH1 T5-3 5.0–5.5 41 97.5 71 31 40 0.750 27.00 5 0.450 CH
BH2 T6-1 7.2–7.7 25 89.1 31 NP NP – – 14 0.415 ML
BH2 T6-2 7.2–7.7 31 96.5 37 21 16 0.375 – 14 0.380 CL
BH2 T6-3 7.2–7.7 50 94.7 68 29 39 0.462 27.40 14 0.410 CH
BH2 T6-4 7.2–7.7 33 99.4 45 25 20 0.600 26.80 14 0.450 CL
BH6 T7-1 7.5–8.0 38 99.4 43 24 19 0.263 – 9 0.400 CL
BH6 T7-2 7.5–8.0 27 86.3 26 NP NP – 26.90 9 0.360 ML
BH8 T9-1 7.5–8.0 48 97.5 68 28 40 0.500 – 7 0.430 CH
BH8 T9-2 7.5–8.0 22 74.1 29 22 7 1.000 – 7 0.380 CL-ML
BH8 T9-3 7.5–8.0 44 99.4 50 21 29 0.207 26.70 7 0.404 CL
BH9 T10-1 9.5–10.0 30 99.7 48 25 23 0.783 – 13 0.318 CL
BH1 T11-1 9.5–10.0 39 100.0 70 26 44 0.705 – 7 0.464 CH
BH1 T11-2 9.5–10.0 43 100.0 56 26 30 0.433 – 7 0.490 CH
wn, water content; FC, fines content; LL, liquid limit; PL, plastic limit; PI, plasticity index; Ic, consistency index; γs, specific gravity; N30, SPT blow count value; and CSR, cyclic
stress ratio.
0
10
20
30
40
50
60
70
80
90
100
0.001
0.01
0.1
1
10
Percent
Finer
Particle Size (mm)
S3-1
S3-2
S3-3
S3-4
S3-5
S3-6
S4-1
S6-2
S7-1
S7-2
S7-3
S8-1
S8-2
S8-3
MH
(PI=31-37)
SM-ML
(PI=NP-20)
Fig. 4. Grain size distribution of test specimens (SM, ML, MH).
0
10
20
30
40
50
60
70
80
90
100
0.001
0.01
0.1
1
10
Percent
Finer
Particle Size (mm)
T3-1
T5-1
T5-2
T5-3
T6-1
T6-2
T6-3
T6-4
T7-1
T7-2
T9-1
T9-2
T9-3
T10-1
T11-1
T11-2
T12-1
T12-2
PI=7
PI=44
Fig. 5. Grain size distribution of test specimens (CL, CH).
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 87

6. the depth of the sample taken was applied as an effective
confining stress for isotropic conditions in a gradually increasing
manner to the specimen.
Test specimens (sandy soils) that had the potential to spread
were firstly frozen and then placed in a test cell. In order to
prevent the specimens from spreading, a vacuum pressure of
10 kPa was applied to the specimens via the pipe at the top cap.
Later, carbon dioxide (CO2) was applied to the soils that are gas
permeable for duration of 1 h. For other soils that did not
permeate gas, this stage was not performed.
An undisturbed specimen of about 50 mm in diameter and
100 mm in height was trimmed and set between the upper and
lower pistons. Next, it was consolidated by isotropic stress for 24 h
under the same effective confining stress (100 kPa). After the
completion of consolidation, a back pressure of 200 kN/m2
was
applied to the sample. The measured B values of the coefficient of
pore water pressure were over 0.95 before cyclic loading. The
details of the cyclic triaxial test series are given in Tables 4 and 5.
For the determination of soil behavior by laboratory testing
under cyclic loading conditions, two different sets of cyclic triaxial
tests were conducted on undisturbed soil samples. The definition of
liquefaction for clayey cohesive soils and sandy or silty soils is given
as the degradation of strength with the number of cycles and with
the corresponding accumulated strain and pore water pressure
ratios reaching ru¼1.0 (¼Δu/σ'c) with the number of cycles,
respectively [26]. The first set of tests, in which cyclic loading had
been applied regarding the failure criterion of the axial strains
reaching ε¼ 72.5% (double-amplitude axial strain attains 5%) or
pore water pressure ratios reaching ru¼1.0 (¼Δu/σ'c), were con-
ducted on non-plastic silty sands or very sandy silts and soils having
a plasticity index at different cyclic stress ratios (CSR¼σd/2σc).
The second set involved tests with a constant frequency applied
(f¼0.1 Hz) for a specific number of cycles, i.e. N¼20 was chosen to
represent the August 17, 1999 earthquake with Mw ¼7.4 at differ-
ent cyclic stress ratios. Just after that, with no change in drainage
and stress conditions, strain-controlled monotonic tests with a
loading speed of 0.20 mm/min were performed at the end of cyclic
loading application stage.
Relatively few equations for the prediction of the number of
cycles of motion have been published. Detailed studies were
performed for the dependency of Ncyc on Mw, Rrup, and the depth
of soil and soil property. Liu et al. [27] developed predictive
models for the median and the aleatory variability of Ncyc as a
function of distance, magnitude, for both deep soil sites and
shallow stiff soil/rock sites. Stafford and Bommer [28] reported
that number of equivalent cycles, NRR(2.0) depends on magnitude
and distance from fault rupture. Haldar and Tang [29] proposed a
relationship between equivalent number of stress cycles, Neq and
earthquake magnitude M expressed in Richter's scale.
According to Liu et al. [27], the equivalent number of cycles,
Ncyc, was approximately determined to be 21 for M¼7.5 and
Rrup E7 km. Number of equivalent cycles was determined as
NRR(2.0)¼23 using the curves which were developed by Stafford
and Bommer [28] depending on Mw and rupture distance. Based
on the approach proposed by Haldar and Tang [29], Neq was
determined approximately to be 18 for M¼7.4. Erken and Ulker
[22], in view of their test results, suggested that N¼20 cycles is
suitable for an earthquake with the magnitude of 7.5. Therefore, in
this study, the choice of Ncyc¼20 cycles corresponding to Mw ¼7.4
earthquake is consistent with the literature.
The frequency value for this study was 0.1 Hz. The earthquake
in Kocaeli lasted approximately 45 s in the outcrop rock zone at
Sakarya Station (Fig. 1. Soil cross section at Sakarya Station). It was
anticipated from measurements made during the earthquake
aftershocks in rock and soft soils that the duration would have
been longer in the soft soil. In this study, the frequency of 0.1 Hz
Fig. 6. Experimental setup of cyclic triaxial test.
Table 4
Dynamic data for the first series tests.
Test no. FC (%) γd
nnn
kN/m³ wn (%) PI (%) wn/wL Soil type B (%) CSR N ε¼ 72.5% Nn
εn
(7%) ru
n
εnn
(7%) ru
nn
S3-1 53 – 37 NP – ML 96 0.280 7 8 8.0 0.98 8 0.98
S3-2 46 – 29 NP – SM 96 0.205 133 152 6.3 1.00 0.16 0.43
S3-3 39 – 32 NP – SM 96 0.210 23 25 7.0 1.03 0.5 0.8
S3-4 92 – 46 22 0.737 MH 96 0.331 27 73 5.9 0.89 2.01 0.51
S3-5 94 – 50 16 0.913 MH 96 0.350 21 58 7.9 0.90 2.47 0.67
S3-6 94 – 46 20 0.786 MH 96 0.377 15 44 11.6 1.00 4.93 0.73
T12-1 98 13.71 33 16 0.876 CL 98 0.346 28 30 2.6 1.00 2.08 0.95
T12-2 95 12.66 33 15 0.995 CL 95 0.315 42 49 2.6 0.95 1.5 0.85
γd, dry unit weight; wn, water content; wL, liquid limit; B, the degree of saturation; CSR, cyclic stress ratio; N, the number of cycles; ε, axial strain; and ru, pore water
pressure ratio.
n
End of the test
nn
Values at the number of cycles, N¼20
nnn
End of the consolidation
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
88

7. was chosen and time was taken as 200 s for the number of cycles
of N¼20. The reason for choosing the frequency value as 0.1 Hz
was because there was no time to measure the pore water
pressure, which might occur in a higher frequency. However, if
higher frequencies were chosen, there would be timing and
continuity problems in measuring the pore water pressure. Sud-
den change in pore water pressure is important. This is supported
in the literature; for example, Zergoun and Vaid [30] demon-
strated that pore pressure measurements were generally not
reliable during cyclic loading tests conducted rapidly (e.g., 1 Hz)
because of the inability of water pressure to equilibrate through-
out the soil specimen and measurement system.
Also, Sancio et al. [20] and Bray and Sancio [12] reported that
“the calculation and comparison of the effective stress path in the
1 Hz and 0.005 Hz cyclic tests provided evidence of the limitations
of the measurements of pore water pressure changes on clayey
silts and clays during a 1 Hz cyclic triaxial test. For this reason, it is
unclear if these soils reach a state of zero effective stress during
undrained cyclic loading at 1 Hz”.
Furthermore, Boulanger and Idriss [31] concluded that the term
slow means that the tests are performed sufficiently slow to
ensure reliable measurements of pore water pressure, as opposed
to the more common seismic loading rate of 1 Hz at which pore
pressure measurements are unreliable for clay samples. The axial
strain and excess pore water pressure responses with loading
cycles representatively are shown in Fig. 7 for non-plastic silt and
clay. It can also be seen that the axial strain and pore water
pressure increased monotonically with the number of loading
cycles. Pore water pressure increases with increasing number of
loading cycles until it becomes equal to the total confining stress,
and axial strain increases with increasing number of loading cycles
reaching up to number of the cycles of N¼20.
At a high CSR of 0.400 and initial effective confining stress of
100 kPa, the silt specimen quickly generates positive pore pressures,
which reach the initial confining pressure in 15 cycles of loading
and reach a double-amplitude strain of 5% in 19 cycles of loading.
At a high CSR of 0.45 and initial effective confining stress of
100 kPa, the clay specimen slowly generates positive pore pres-
sures, which are approximately constant after 16 cycles of loading
and reach a double-amplitude strain of 5% in 5.5 cycles of loading.
The static tests were terminated at the displacement value
corresponding to 10% of axial strain because of the limitation of
the cyclic triaxial testing apparatus (the maximum value of axial
strain measured is 720 mm).
The S8-3 test sample, which was taken close to the surface
(3.0–3.5 m), was non-plastic, low plasticity silt classified as ML.
The water content of this sample was wn ¼44%, the fines content
was FC¼97.3%, and the plasticity index was non-plastic, PI¼0.
The T5-3 test sample, which was taken close to the surface
(5.0–5.5 m), was high plasticity clay classified as CH. The water
content of this sample was wn ¼41%, its fines content was
FC¼97.5%, and the plasticity index was PI¼40.
In the clay sample subjected to a cyclic stress ratio of CSR¼0.450,
pore pressures accumulated rapidly during the first 5 cycles and the
sample experienced a large cyclic axial strain at the end of 20 cycles.
Although the pore pressure ratio was also obtained for the S8-3 test
sample with ML and PI¼NP, the PWP ratio did not reach the value of
ru¼1.0 for CH soil with a plasticity index of PI¼40. However, axial
strains remained limited due to the overconsolidation of surface soils.
The results obtained for CH soils are consistent with pore water
pressure behavior in the laboratory as described by Zehtab [13].
6. Cyclic stress–strain-pore pressure behavior of undisturbed
soils
Soils were considered in two separate groups as silty sands (sand-
like soils) and fine-grained soils (clay-like soils). The stress–strain
Table 5
Dynamic data for the second series tests.
Test no. Borehole no. FC (%) End of cons. γd (kN/m³) PI (%) Soil type B (%) σc (kPa) CSR N ε¼ 72.5 (%) At the N¼20 σds ε¼10% (kPa)
ε (7%) ru
S4-1 BH4 56 17.00 – ML 96 100 0.320 19.3 2.8 0.96 187
S6-1 BH6 97 14.20 23 CH 100 100 0.420 – 2.1 1.00 129
S6-2 BH6 99 12.80 38 MH 96 100 0.482 18.3 2.7 1.00 121
S7-1 BH7 98 11.80 33 MH 92 100 0.370 – 0.9 0.48 127
S7-2 BH7 97 11.60 36 MH 92 100 0.420 – 1.7 0.29 122
S7-3 BH7 99 12.90 31 MH 94 100 0.515 15.3 2.8 1.00 188
S8-1 BH8 87 14.10 NP ML 98 100 0.240 – 0.7 0.54 149
S8-2 BH8 89 14.10 NP ML 94 100 0.255 – 1.2 0.81 139
S8-3 BH8 97 13.50 NP ML 94 100 0.400 19.2 2.9 1.00 120
T3-1 BH4 97/9 18.72 16/NP CL/SP-SC 96 100 0.320 – 0.4 0.78 315
T4-1 BH2 1 15.76 NP SP 100 100 0.408 – 4.8 1.00 565
T5-1 BH1 99 17.55 40 CH 96 100 0.400 – 1.34 0.90 128
T5-2 BH1 98 18.42 16 CL 94 100 0.340 – 0.51 0.77 182
T5-3 BH1 97 17.20 40 CH 99 100 0.450 5.5 6.1 0.85 93
T6-1 BH2 89 17.63 NP ML 96 100 0.415 20 2.65 0.98 282
T6-2 BH2 96 18.32 16 CL 96 100 0.380 – 1.11 0.83 228
T6-3 BH2 95 16.79 39 CH 98 100 0.410 – 1.36 0.86 94
T6-4 BH2 99 17.76 20 CL 96 100 0.450 1 10.7 0.80 34
T7-1 BH6 99 17.79 19 CL 100 100 0.400 – 1.63 0.93 154
T7-2 BH6 86 17.15 NP ML 96 100 0.360 2.5 4.90 1.00 328
T9-1 BH8 97 16.67 40 CH 94 100 0.430 7.5 4.15 0.71 109
T9-2 BH8 74 18.96 7 CL-ML 94 100 0.380 – 1.55 0.81 349
T9-3 BH8 99 18.07 29 CL 94 100 0.404 – 1.47 0.95 142
T10-1 BH9 100 17.91 23 CL 94 100 0.318 – 0.68 0.66 164
T11-1 BH1 100 17.76 44 CH 96 100 0.464 – 1.30 0.88 127
T11-2 BH1 100 17.83 30 CH 96 100 0.490 20 2.42 1.00 157
T12-1 BH3 98 13.71 16 CL 98 100 0.346 28 1.97 0.95 206
T12-2 BH3 95 12.66 15 CL 95 100 0.315 42 1.50 0.85 160
FC, fines content; PI, plasticity index; B, the degree of saturation; σc, confining pressure; CSR, cyclic stress ratio; N, the number of cycles; ε, axial strain; ru, pore water pressure
ratio; and σds deviator stress for ε¼10%.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 89

8. behavior of soils was investigated by taking into account such
parameters as wn/wL, which provide information on the consistency,
saturation degrees (B), fines content (FC), and overconsolidation
ratio (OCR).
6.1. Stress–strain behaviors of silty sands
The effects of cyclic stress ratio on the behavior of axial strain
and excess pore water pressure versus number of cycles are shown
in Fig. 8 for non-plastic silty sand and silt soils with fines content
varying from 39% to 53%.
As can be observed from the response patterns given in Fig. 10,
the increased cyclic shear stress ratio causes it to reach the failure
limit of double-amplitude shear strain of 5% in lower cycles of
loading. Although the cyclic stress ratios of the S3-2 and S3-3
samples were approximately the same, the S3-2 sample had 7%
more fines content than the S3-3 sample, and it reached the value
of ru ¼1 in 146 cycles of loading.
One of the important parameters controlling the cyclic stress–
strain characteristics of soils is the number of cycles. This para-
meter plays a crucial role especially in analyzing the behavior of
soil layers under earthquake loads. As shown in Fig. 8, in the S3-1
soil sample subjected to cyclic shear stress ratio of CSR¼0.280,
axial strain rapidly reached the value of ε¼ 72.5% in N¼8 cycles
of loading.
6.2. Stress–strain behaviors of fine-grained soils
The effect of the plasticity index on the behavior of axial strain
and pore pressure versus number of cycles of soils at almost the
same stress ratio are shown in Figs. 9 and 10. The effect of
plasticity index on the behavior of stress–strain of non-plastic silt
and low-medium plastic clay with a stress ratio in the range of
0.360–0.380 is shown in Fig. 9.
As can be observed from Fig. 9, the largest axial strain
(ε¼ 75%) and pore water pressure ratio (ru ¼1.0) occurred in the
non-plastic silt sample (T7-2). Axial strains and pore water
pressures at the end of testing were reduced by increasing the
plasticity index. The pore water pressure value of the T6-2 test
sample (PI¼16) was lower than that of the T9-2 test sample due to
low plasticity. Also, due to high plasticity, the lack of complete
saturation and stiff consistency (wn/wL ¼0.554) in the S7-1 sample,
which was slightly overconsolidated due to its being a surface
sample obtained from 3.0 m depth, the axial strains and pore
water pressures remained limited (Fig. 9).
The variation in stress–strain behavior of the soils close to the
surface (3.5–5.0 m), with a plasticity index of between NP and 40,
is given in Fig. 10 for stress ratios ranging from 0.400 to 0.420.
The applied cyclic axial stress levels were approximately the
same but the plasticity index changed within a large range of NP to
40% (Fig. 10). As can be observed from the response patterns given
in Fig. 10, more axial strain and pore water pressure occurred in
-0.5
-0.4
-0.3
-0.2
-0.1
0.0
0.1
0.2
0.3
0.4
0.5
σ
σ
ε
0 2 4 6 8 10 12 14 16 18 20 22
Stress
Ratio,
d
/2
c
-0.5
-0.4
-0.3
-0.2
-0.1
0.0
0.1
0.2
0.3
0.4
0.5
0 2 4 6 8 10 12 14 16 18 20 22
-4.0
-3.0
-2.0
-1.0
0.0
1.0
2.0
3.0
4.0
0 2 4 6 8 10 12 14 16 18 20 22
Axial
Strain,
(%)
-8.0
-6.0
-4.0
-2.0
0.0
2.0
4.0
6.0
8.0
0 2 4 6 8 10 12 14 16 18 20 22
-0.4
-0.2
0.0
0.2
0.4
0.6
0.8
1.0
0 2 4 6 8 10 12 14 16 18 20 22
Pwp
Ratio,
r
u
Number of Cycles, N
-0.4
-0.2
0.0
0.2
0.4
0.6
0.8
1.0
0 2 4 6 8 10 12 14 16 18 20 22
Number of Cycles, N
Fig. 7. Results of the cyclic triaxial tests on the S8-3 (left hand side) and the T5-3 (right hand side) samples in which properties are given in Tables 2 and 3: (a) deviator stress
versus number of cycles; (b) axial strain versus number of cycles; and (c) excess pore water pressure versus number of cycles.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
90

9. the fine-grained soils with low plasticity index. As the pore water
pressure value of the S7-2 test sample was expected to be between
those of the S6-1 and T5-1 test samples, this value is low. The
reason for this is the lower saturation degree of B¼92% and
consistency of wn/wL¼0.556 in the S7-2 sample. Although the
pore pressure ratio of ru ¼1.0 was also obtained for the S6-1 test
sample with CH and PI¼23, the PWP ratio did not reach the value
of ru ¼1.0 for MH and CH soil groups with a plasticity index value
greater than 23%.
6.3. Plasticity index effect on axial strain
In this part of the study, the variations in axial strains which
occurred as a result of cyclic loading depending on plasticity index
were investigated in surface soils (3.0–5.5 m) and in deeper soil
layers (7.7–10.0 m) under different cyclic stress ratios between
0.240 and 0.450 (Fig. 11).
Axial strain decreases with increasing plasticity index due to
cohesive force. Increases in the cyclic shear stress ratio cause an
increase in axial strains in every plasticity index. However, axial
strains remained limited due to the overconsolidation of surface
soils. Also, samples obtained from the depth of 3.0 m by Sancio
[21] with a cyclic stress ratio of 0.405 are shown in Fig. 11.
6.4. Plasticity index effect on pore water pressure
The relationship between pore water pressure and plasticity
was redrawn according to different axial shear strain levels for
4 chosen cyclic stress ratios (CSR¼0.35–0.40–0.45–0.50) and
plasticity indices (PI¼NP-10–20–30–40). Graphics were produced
taking into account critical shear strain at the axial shear strain of
0
2
4
6
8
Axial
Strain,
±%
Number of Cycles, N
S3-3, CSR=0.210, PI=NP, FC=39%
S3-2, CSR=0.205, PI=NP, FC=46%
S3-1, CSR=0.280, PI=NP, FC=53%
0.0
0.2
0.4
0.6
0.8
1.0
0 20 40 60 80 100 120 140 160
0 20 40 60 80 100 120 140 160
Pore
Pressure
Ratio
Number of Cycles, N
S3-3, CSR=0.210, PI=NP, FC=39%
S3-2, CSR=0.205, PI=NP, FC=46%
S3-1, CSR=0.280, PI=NP, FC=53%
Fig. 8. Stress–strain behavior of samples, which have CSR values ranging from
0.205 to 0.280.
0
2
4
6
0 5 10 15 20 25
Axial
Strain,
±%
Number of Cycles, N
T7-2, CSR=0.360, PI=NP, FC=86%
T9-2, CSR=0.380, PI=7, FC=74%
T6-2, CSR=0.380, PI=16, FC=96%
S7-1, CSR=0.370, PI=33, FC=98%
0.0
0.2
0.4
0.6
0.8
1.0
0 5 10 15 20 25
Pore
Pressure
Ratio
Number of Cycles, N
T7-2, CSR=0.360, PI=NP, FC=86%
T9-2, CSR=0.380, PI=7, FC=74%
T6-2, CSR=0.380, PI=16, FC=96%
S7-1, CSR=0.370, PI=33, FC=98%
Fig. 9. Stress–strain behavior of samples, which have CSR values ranging from
0.360 to 0.380.
0
1
2
3
Axial
Strain,
±%
Number of Cycles, N
S8-3, CSR=0.400, ML, PI=NP
T7-1, CSR=0.400, CL, PI=19
S6-1, CSR=0.420, CH, PI=23
S7-2, CSR=0.420, MH, PI=36
T6-3, CSR=0.410, CH, PI=39
T5-1, CSR=0.400, CH, PI=40
0.0
0.2
0.4
0.6
0.8
1.0
0 5 10 15 20 25
0 5 10 15 20 25
Pore
Pressure
Ratio
Number of Cycles, N
S8-3, CSR=0.400, ML, PI=NP
T7-1, CSR=0.400, CL, PI=19
S6-1, CSR=0.420, CH, PI=23
S7-2, CSR=0.420, MH, PI=36
T6-3, CSR=0.410, CH, PI=39
T5-1, CSR=0.400, CH, PI=40
Fig. 10. Stress–strain behavior of samples, which have CSR values ranging from
0.400 to 0.420.
0
1
2
3
4
ε
0 10 20 30 40 50
Axial
Strain,
(±%)
Plasticity Index, PI
CSR=0.240-0.255
CSR=0.315-0.330
CSR=0.346-0.370
CSR=0.400-0.450
Sancio, 2003
Fig. 11. Relationship between axial strain and plasticity index at the end of 20
cyclic loading application stages (depth is in the range from 3.0 m to 5.5 m).
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 91

10. ε¼ 70.75–1.5–2.5% for resultant pore water pressure according to
the chosen plasticity index.
The relationship between CSR and PWP ratio is given in
Figs. 12–14 for different strain levels depending on plasticity.
Induced pore water pressure is affected by the plasticity index
and particle size and properties. If soil is non-plastic silt, pore
water pressure increases even at low strain level. In Fig. 12, the
PWP ratio did not reach the value of 1 due to the lower CSR at the
axial shear strain of ε¼ 70.75% for non-plastic soils. In this figure,
the PWP ratio is ru ¼1.0 for the plasticity index of PI¼10–20 and
CSR of σd/2σc¼0.500 or greater. The PWP ratio remained at the
levels of ru ¼0.80 and 0.55, respectively, for plasticity index of
PI¼30–40 at maximum CSR level. As shown in Fig. 12, PWP
becomes ru ¼0.85, even though CSR is as low as 0.205. ru will
reach a value of 1.0 rapidly when CSR is greater than 0.205.
The relationship between CSR and PWP ratio is given in Fig. 13
depending on the plasticity index of PI¼NP and 40 for the axial
shear strain of ε¼ 70.75–1.5–2.5%. The PWP is about 0.13, 0.22
and 0.48 at the axial shear strain of ε¼ 70.75–1.5–2.5%, respec-
tively, when the CSR for high plasticity soils with PI¼40 is 0.40. It
is seen from this figure that PWP increases as axial strain becomes
ε¼ 70.75%. When CSR equals 0.45 for non-plastic soils, the PWP
ratio becomes 1. It is about 0.80 when the CSR for high plasticity
soils with PI¼40 is 0.500. In the same graphics, PWP is about 0.50,
0.75 and 0.90 at the axial shear strain of ε¼ 70.75–1.5–2.5%,
respectively, when the CSR for non-plastic soils is 0.25.
In Fig. 14, at a level of ε¼ 72.5%, which is the failure criterion,
the development of PWP is displayed depending on plasticity
index and CSR. For non-plastic soils, ru equals 1.0 when the CSR
is equal to or greater than 0.45, for soils with plasticity index of
PI¼10–20–30, ru equals 1.0 when the CSR is equal to or greater
than 0.50.
6.5. Dynamic strength
The first set of tests were conducted on non-plastic silty sands
or very sandy silts and soils with plasticity index values of PI¼
15–22, respectively. Information about the first series of tests,
which were performed at different stress ratios using the physical
properties of soil samples given in Table 2, is presented in Table 4.
Fig. 15 shows the relation between the number of cycles causing
5% double-amplitude strain cyclic stress ratio. It is seen in this
figure that the cyclic strength decreased generally as the number
of cycles increased. According to the strength curve obtained by
using the results of these tests, the cyclic strength of silty sands-
sands (FC¼39–53%) is 45% less than the cyclic strength of high
plasticity soils (MH, PI¼15–22%). It was also determined that the
cyclic strength of soils with a plasticity index of PI¼15–16% was
less than that of high plasticity soils.
The results of the second set of tests indicate that the strength
curves belonging to the low-plasticity silts (ML) and clays (CL)
together with the high-plasticity silts (MH) and clays (CH) take
place above the strength curve of silty sands (SM) on the graph of
number of cycles versus cyclic stress ratio. In other words, the
strength increases.
6.6. Effects of soil plasticity
The undrained cyclic stress–strain responses of non-plastic
(S4-1 and T7-2) and plastic (S3-4 and S3-6) specimens, as
compared in Fig. 16, exhibit the characteristics expected from
y = 0.43x0.821
R2 = 0.84
y = 0.495x0.221
R2 = 0.98
y = 0.505x0.167
R2 = 0.99
y = 0.52x0.137
R2 = 0.99
y = 0.55x0.139
R2 = 0.97
y = 0.27x1.3082
R² = 1
0
0.1
0.2
0.3
0.4
0.5
0.6
0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6
Cyclic
Stress
Ratio
Pore Pressure Ratio
PI=0
PI=10%
PI=20%
PI=30%
PI=40%
Sand
Fig. 12. Relationship between cyclic stress ratio and pore pressure ratio based on
plasticity index corresponding to ε¼ 70.75% in the cyclic test.
Fig. 13. Relationship between cyclic stress ratio and pore pressure ratio for high
plasticity soils with PI¼40 and non-plastic soils corresponding to ε¼ 70.75–1.50–
2.5% in the cyclic test.
0
0.1
0.2
0.3
0.4
0.5
0.6
0.0 0.2 0.4 0.6 0.8 1.0 1.2
Cyclic
Stress
Ratio
Pore Pressure Ratio
Fig. 14. Relationship between cyclic stress ratio and pore pressure ratio based on
plasticity index corresponding to ε¼ 72.50% in the cyclic test.
0.0
0.1
0.2
0.3
0.4
0.5
0.6
1000
100
10
1
Cyclic
Stress
Ratio,
/2σ
σ
Number of Cycles, N
MH PI=27-31
CH PI=38-40
MH PI=15-22
CL PI=15-20
ML PI=NP
SM FC=39-46%
Fig. 15. Dynamic strength curves of the samples used in this study.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
92

11. sand-like and clay-like soils, respectively. S3-4 and S3-6, with their
higher plasticity, had broader hysteretic loops.
Maximum deviator stress of 62 kPa was applied to the samples
given in Fig. 16a. More axial shear strains were observed in the
region of compression for the S4-1 and S3-4 samples. An inclined
loop of the non-plastic soil including FC¼56% was higher than that
of the sample with plastic characteristics and a high fines content.
In Fig. 16, it can be seen that the energy dissipation property (loop
area) of the soil sample with a higher plasticity (S3-4) is more than
about 20%. However, as the plasticity index increases, the number
of cycles required to reach the failure limit (ε¼ 72.5%) increases.
Maximum deviator stress of 70 kPa was applied to the samples
given in Fig. 16b. More axial shear strains occurred in the
extension zone of the S3-6 and T7-2 samples. The behavior of
the loops is similar to that in Fig. 16a. The number of cycles
required to reach the failure limit for the T7-2 sample was less
than for the S3-6 sample with a high plasticity. The T7-2 sample is
non-plastic soil (ML) with a value of wn/wL¼1.038.
The fine-grained soils with different plasticity used in this
study are illustrated in the plasticity chart showing the recom-
mendations of Boulanger and Idriss [32] (see Fig. 17). They
explained the cyclic behavior of soils with the term “sand-like”
for liquefaction, and with the term “clay-like” for cyclic failure.
They also suggest an intermediate behavior for soils classified as
CL–ML and ML with PI values of 4–7.
Sand-like behavior was observed for silty sand, low plasticity
silt samples and high plasticity silt samples with a B value greater
than 95%. Clay-like behavior was observed for soils classified as CL,
CH and MH with a B value less than 92%. The undisturbed soil
specimen with PI¼7 (T9-2) presented intermediate behavior
which is found to be in line with the results of Boulanger and
Idriss [32].
-80
-60
-40
-20
0
20
40
60
80
-4 -3 -2 -1 0 1 2 3 4
Deviator
Stress
(kPa)
S4-1, N=1
S4-1, N=19
-80
-60
-40
-20
0
20
40
60
80
-4 -3 -2 -1 0 1 2 3 4
S3-4, N=1
S3-4, N=27
-80
-60
-40
-20
0
20
40
60
80
-4 -3 -2 -1 0 1 2 3 4
Deviator
Stress
(kPa)
Axial Strain (%)
T7-2, N=1
T7-2, N=2.5
-80
-60
-40
-20
0
20
40
60
80
-4 -3 -2 -1 0 1 2 3 4
Axial Strain (%)
S3-6, N=1
S3-6, N=15
Fig. 16. Stress–strain relationship for the first cycle of loading (fine line) and the cycle at which 72.5% axial strain or ru¼1.0 is reached (thick line) for four specimens of
increasing plasticity. The tests were performed on soils initially consolidated to an effective stress of 100 kPa.
0
10
20
30
40
50
0 10 20 30 40 50 60 70 80
Plasticity
Index,
PI
(%)
Liquid Limit, LL (%)
Clay-like behavior (Boulanger et al., 2004)
Transition of sand-like to clay-like behavior (Boulanger et al., 2004)
Sand-like behavior (Boulanger et al., 2004)
Present Study (UD)
Fig. 17. Plasticity chart showing fine grained soils that exhibit clay-like, inter-
mediate and sand-like behaviors that were used at the cyclic triaxial tests.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 93

12. The effects of fine grained soils with different fines content and
plasticity on dynamic behaviors are observed to be inconsistent.
The plasticity index can rightly recognize fine-grained soils'
behavior either “clay-like” which is expected to have cyclic soft-
ening or as “sand-like” which is liquefiable. Therefore, for both
dynamic behavior and bearing capacity loss determination, it
becomes very important to utilize laboratory cyclic tests to
eliminate the influence of such factors.
6.7. Determination of critical deformation level
Because of soil softening, the axial strain increased when
dynamic shear stress has sufficiently high magnitude to constitute
axial strain and pore water pressure. There is a critical level of
deformation when the effective stress decreases along with the
rapidly increasing axial shear strain. We aimed to define the level
of the critical deformation for each soil specimen in the scope of
this research. Therefore, the relationship between mean effective
stress and deformation was examined. Effective stress is defined as
σ0
c ¼ σc Δu ð1Þ
where σc is consolidation pressure (σc¼100 kPa) and Δu is the
excess pore water pressure during the dynamic loading.
The mean effective stress (p') is calculated by the following
formula:
p0
¼
σ0
1 þσ0
2 þσ0
3
3
ð2Þ
The relationship between mean effective stress and axial shear
strain is shown in Fig. 18. As indicated in the figure, the level of
critical deformation is defined as the point of intersection of the
tangents.
The critical level of deformation for the specimens of silty sand
(SM) and low plasticity normally consolidated silt is εcritic¼ 70.50%,
as shown in Fig. 18. The axial shear strains increase rapidly and
exceed the critical failure level (ε¼ 72.50%) after this level of
critical deformation. The obtained critical deformation levels,
εcritic¼ 70.50%, are compatible with the level of deformation stated
in the study conducted by Ansal and Erken [33] for kaolin clay with
PI¼40% and in the study conducted by Erken and Ulker [22] for silt
with PI¼5%. The critical deformation level of soils with plasticity
index values of PI¼15–16 and overconsolidation ratio (OCR) values
ranging from 2.75 to 3.85, is determined as εcritic¼ 71.0% in Fig. 19.
7. Results of the static tests after dynamic tests
The second set involved the dynamic tests for number of cycles
of N¼20 at different cyclic stress ratios and just after that, with no
change in drainage and stress conditions, the strain-controlled
monotonic tests (0.20 mm/min) were performed. Deviator stress
(σds) corresponding to an axial strain of 10% or failure point was
found as a result of the static test. The results of static tests
performed after dynamic tests are given in Table 5.
7.1. Plasticity index effect on post-dynamic behavior
The behavior of soils, which have a CSR of 0.42 and two soil
groups (MH and CH) with high plasticity as the result of the
dynamic loading and static test performed after dynamic loading,
is shown in Fig. 20, which depicts the relationship between shear
stress and pore water pressures based on axial shear strain. As can
be seen in Fig. 20, the shear strains and pore water pressures
increase steadily with the increasing number of cycles. Therefore,
the inclination of loops will decrease toward the x-axis due to
strain softening. The axial strain level and pore water pressure
following a cyclic shear stress level application of σd ¼ 785.0 kPa
were approximately γc¼ 72.06–1.36% and 100–87 kPa at these
strain levels for S6-1 and T6-3, respectively.
When maximum deviator stress reached nearly 151 kPa for the
S6-1 test sample in the static test, the pore water pressure
decreased to 52 kPa. However, while the maximum deviator stress
was obtained as 87 kPa for the T6-3 test sample, the pore water
pressure decreased to 45 kPa at the end of the test (Fig. 20).
-8
-4
0
4
8
0 20 40 60 80 100 120
Axial
Shear
Strain
(±%)
Mean Effective Stress, p' (kPa)
Fig. 18. Axial strain–effective stress behavior of samples exhibiting sand-like
behavior under dynamic loading.
-4
-2
0 20 40 60 80 100 120
Axial
Shear
Strain
(±%)
Mean Effective Stress, p' (kPa)
Fig. 19. Axial strain–effective stress behavior of samples exhibiting clay-like
behavior under dynamic loading.
-100
-50
0
50
100
150
200
-4 -2 0 2 4 6 8 10 12 14
Deviator
Stress
(kPa)
Axial Strain (%)
-60
-40
-20
0
20
40
60
80
100
120
-4 -2 0 2 4 6 8 10 12 14
Pore
Water
Pressure
(kPa)
Axial Strain (%)
Fig. 20. The variation among the deviator stress and pore water pressure versus
axial strain of the S6-1 and the T6-3 test samples.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
94

13. The effects of plasticity on the behavior of soils of low plasticity
silt and high plasticity clay are shown in Fig. 21. While axial shear
strain and pore water pressure were measured as 72.65% and
98 kPa for the non-plastic T6-1 test sample at the end of 20 cycles,
these values were obtained as 71.36% and 85 kPa for the T6-3 test
sample, respectively. The post-cyclic ultimate monotonic static
deviator stress value was obtained as 280 kPa at 10% axial shear
strain for the T6-1 test sample. This value was 94 kPa for the T6-3
sample. As can be observed in Fig. 21, as failure had been expected
at low deviator stress in the static loading process for the T6-3 test
sample, which was subjected to high axial strain during the cyclic
loading, there was a greater stress value. The high deviator stress
was provided by the negative pore water pressure resulting from
dilation of the sample in the static loading stage.
7.2. Post-cyclic shear strength
The results of the conducted dynamic and subsequent static
tests are given in Table 5. The reduction in static strength due to
repeated loading is depicted by taking into account the cyclic
strength curves (Fig. 15) and the behavior of the cyclic stress–
strain in the study.
The relationship between axial strain (εc) caused by cyclic
loading at different cyclic stress ratios for N¼20 and deviator
stress corresponding to axial strain of 10% as a result of the static
test after cyclic loading are shown in Fig. 22. In this figure, the
reduction in deviator stress according to soil groups is shown.
Because high negative pore pressure occurred depending on the
level of deformation for non-plastic soils, greater deviator stress
was obtained for these soils. However, deviator stress did not
change much in high plasticity soils due to non-formation of
excess negative pore water pressure with increased plasticity
index. The reductions in the deviator stresses obtained in the
cases of static test after cyclic loading tests at different cyclic stress
ratios for N¼20 were 27%, 40%, 53%, and 65% for soils with
plasticity indices of NP, PI¼16%, 30% and 40%, respectively.
If deformations occurring during earthquakes are equal to or
less than the critical deformation level, the shear strength losses of
soil are not significant. If deformations occurring during earth-
quakes increase, the post-cyclic strength of fine-grained soils
decreases depending on the deformation levels.
8. Conclusions
To study the mechanism of softening of fine grain soils under-
lying in Adapazari city, a series of cyclic triaxial tests and post-
cyclic monotonic tests have been conducted. The main conclusions
are as follows:
1. Both the axial strain and excess pore water pressure increased
with increasing number of loading cycles in samples. The cyclic
strength tended to decrease as the amplitude of the cyclic stress
ratio increased. The cyclic strength (number of cycles causing 5%
double-amplitude strain or reaching ru¼1.0 pore water pressure
ratios) increased with increasing plasticity index.
2. The dynamic strength of silty sand (FC¼39–46%) was found to
be 45% lower than that of high plasticity soils (MH) with
PI¼15–22%. The soils with a plasticity index of PI¼15–16%
had lower strength compared to high plasticity soils.
3. It was found that silty sand soils had the lowest strength. Also,
the dynamic strength of the soils increased along with the
increase in plasticity.
4. Sand-like behavior was observed for silty sand, low plasticity
silt and high plasticity silt samples with a B value greater than
95%. Clay-like behavior was observed for soils classified as CL,
CH and MH with a B value less than 92%. The undisturbed soil
specimen with PI¼7 (T9-2) presented intermediate behavior.
5. The experimental results showed that induced pore water
pressure is affected by the plasticity index, particle size and
properties. If soil is non-plastic silt pore water, pressure
increases even at low strain level.
6. The critical level of the deformation for the specimens of the
silty sand (SM) and low plasticity normally consolidated silt
was found as εcritic ¼ 70.50%. The critical deformation level of
the oversconsolidated soils was determined as εcritic ¼ 71.0%.
7. The static strength of the soils exposed to the excessive axial
strain during the cyclic loading decreased after cyclic loading.
The reductions in the deviator stresses obtained in the cases of
static test after cyclic loading tests at different cyclic stress
ratios for a number of cycles of N¼20 were 27%, 40%, 53%, and
65% for soils with plasticity indices of NP, PI¼16%, PI¼30% and
PI¼40%, respectively.
-150
-100
-50
0
50
100
150
200
250
300
350
400
Deviator
Stress
(kPa)
Axial Strain (%)
-60
-40
-20
0
20
40
60
80
100
120
-5 0 5 10 15 20 25
Pore
Water
Pressure
(kPa)
Axial Strain (%)
Fig. 21. The variation among the deviator stress and pore water pressure versus
axial strain of the T6-1 and the T6-3 test samples.
0
50
100
150
200
250
300
350
400
0
0
1
0
1
1
1
.
0
Static
Strength
(Deviator
Stress
at
the
failure)
(kPa)
Strain at the end of Dynamic Test, εc (±%)
Fig. 22. Relationship between axial strain at the end of the dynamic test and
deviator stress corresponding to ε¼10% in the monotonic test.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96 95

14. 8. It is concluded that clayey soils with high plasticity (PIr30)
and low consistency index (Ic r0.433) exhibited softening
behavior at the end of the number of cycles of N¼20.
Acknowledgments
The authors would like to thank Japan International Coopera-
tion Agency for the financial support provided by the Turkey–
Japan Earthquake Disaster Prevention Center Project (1993–2002).
References
[1] Barka A, Akyuz S, Altunel E, Sunal G, Cakir Z, Dikbas A, et al. August 17, 1999
Izmit earthquake, M¼7.4, Eastern Marmara region, Turkey: study of surface
rupture and slip distribution. The 1999 Izmit and Duzce earthquakes:
preliminary results. Istanbul Technical University; 2000. p. 15–30.
[2] Sancio RB, Bray JD, Stewart JP, Youd TL, Durgunoglu HT, Onalp A, et al.
Correlation between ground failure and soil conditions in Adapazari, Turkey.
Soil Dyn Earthq Eng 2002;22:1093–102.
[3] Bakır BS, Sucuoglu H, Yılmaz T. An overview of local site effects and the
associated building damage in Adapazari during the 17 August 1999 Izmit
earthquake. Bull Seismol Soc Am 2002;92:509–26.
[4] Beyen K, Erdik M. Two-dimensional nonlinear site response analysis of
Adapazari plain and predictions inferred from aftershocks of the Kocaeli
earthquake of 17 August 1999. Soil Dyn Earthq Eng 2004;24:261–79.
[5] Mollamahmutoglu M, Kayabali K, Beyaz T, Kolay E. Liquefaction-related
building damage in Adapazari during the Turkey earthquake of August 17,
1999. Eng Geol 2003;67:297–307.
[6] Komazawa M, Morikawa H, Nakamura K, Akamatsu J, Nishimura K, Sawada S,
et al. Bedrock structure in Adapazari, Turkey – a possible cause of severe
damage by the 1999 Kocaeli earthquake. Soil Dyn Earthq Eng 2002;22:829–36.
[7] Rathje EM, Karatas I, Wright SG, Bachhuber J. Coastal failures during the 1999
Kocaeli earthquake in Turkey. Soil Dyn Earthq Eng 2004;24(9):699–712.
[8] Bray JD, Sancio RB, Durgunoglu HT, Onalp A, Youd TL, Stewart JP, et al.
Subsurface characterization at ground failure sites in Adapazari, Turkey.
J Geotech Geoenviron Eng 2004;130(7):673–85.
[9] Cetin KO, Youd TL, Seed RB, Bray JD, Sancio R, Lettis W, et al. Liquefaction-
induced ground deformations at Hotel Sapanca during Kocaeli (Izmit), Turkey
earthquake. Soil Dyn Earthq Eng 2002;22:1083–92.
[10] Cetin KO, Isik N, Unutmaz B. Seismically induced landslide at Degirmendere
Nose, Izmit Bay during Kocaeli (Izmit) – Turkey earthquake. Soil Dyn Earthq
Eng 2004;24:189–97.
[11] Yılmaz MT, Pekcan O, Bakır BS. Undrained cyclic shear and deformation
behavior of silt–clay mixtures of Adapazarı, Turkey. Soil Dyn Earthq Eng
2004;24:497–507.
[12] Bray JD, Sancio RB. Assesment of the liquefaction susceptibility of fine-grained
soils. J Geotech Geoenviron Eng 2006;132(9):1165–77.
[13] Zehtab KH. An assessment of the dynamic properties of Adapazari soils by
cyclic direct simple shear tests. Turkey: METU, Department of Engineering
Sciences, Ankara; 2010 M.S. thesis.
[14] Yılmaz MT, Zehtab KH. Practical method for utilization of commercial cyclic
testing apparatuses for computation of site response in central Adapazari. Soil
Dyn Earthq Eng 2014;63:203–16.
[15] Onalp A, Arel E, Bol E. A general assessment of the effects of earthquake on the
soil–structure interaction in Adapazari 1999. Istanbul, Turkey: XV ICSMFE;
2001. p. 76–89.
[16] ASTM, Standard practice for determining the normalized penetration resis-
tance of sands for evaluation of liquefaction potential. ASTM D 6066-98.
Annual book of ASTM standards 04.09; 2000.
[17] ASTM, Standard test method for penetration test and split barrel sampling of
soils. ASTM D 1589-99. Annual book of ASTM standards 04.09; 2000.
[18] Youd TL, Idriss IM, Andrus RD, Arango I, Castro G, Christian JT, et al.
Liquefaction resistance of soils: summary report from the 1996 NCEER and
1998 NCEER/NSF workshops on evaluation of liquefaction resistance of soils.
J Geotech Geoenviron Eng 2001;127(10):817–33.
[19] JGS, Preparation of soil specimens for triaxial tests. JGS-0520-2000.
[20] Sancio RB, Bray JD, Riemer MF, Durgunoglu T. An assessment of the liquefac-
tion susceptibility of Adapazari silt. In: Proceedings of the 2003 Pacific
Conference on Earthquake Engineering. Christchurch, New Zealand; 13–15
February 2003a. Paper No. 172.
[21] Sancio RB. Ground failure and building performance in Adapazari, Turkey. Ph.
D. thesis. Berkeley, California, USA: University of California; 2003.
[22] Erken A, Ulker MBC. Effect of cyclic loading on monotonic shear strength of
fine-grained soils. Eng Geol 2007;89:243–57.
[23] Thammathiwat A, Chim-oye W. Behavior of strength and pore pressure of soft
bangkok clay under cyclic loading. Thammasat Int J Sci Tech 2004;9(4):21–8.
[24] JGS, Method for cyclic undrained triaxial test on soils. JGS-0541-2000.
[25] JGS, Method for cyclic triaxial test to determine deformation properties of
geomaterials. JGS 0542-2000.
[26] Ishihara K. Soil behaviour in earthquake geotechnics. 1st ed.. Oxford: Claredon
Press; 1996.
[27] Liu AH, Stewart JP, Abrahamson NA, Moriwaki Y. Equivalent number of
uniform stress cycles for soil liquefaction analysis. J Geotech Geoenviron Eng
2001;127(12):1017–26.
[28] Stafford PJ, Bommer JJ. Empirical equations for the prediction of the equivalent
number of cycles of earthquake ground motion. Soil Dyn Earthq Eng
2009;29:1425–36.
[29] Haldar A, Tang WH. Statistical study of uniform cycles in earthquakes.
J Geotech Eng Div 1981;107(GT5):577–89 [Proceeding Paper 16239, ASCE].
[30] Zergoun M, Vaid YP. Effective stress response of clay to undrained cyclic
loading. Can Geotech J 1994;315:714–27.
[31] Boulanger R, Idriss I. Liquefaction susceptibility criteria for silts and clays.
J Geotech Geoenviron Eng 2006;132(11):1413–26.
[32] Boulanger RW, Idriss IM. Evaluating the potential for liquefaction or cyclic
failure of silts and clays, Report No. UCD/CGM-04/01, Center for Geotechnical
Modeling. Davis, California: Department of Civil and Environmental Engineer-
ing University of California; 2004.
[33] Ansal AM, Erken A. Undrained behavior of a clay under cyclic shear stresses.
J Geotech Eng Div 1989;115(7):968–83.
Z. Kaya, A. Erken / Soil Dynamics and Earthquake Engineering 77 (2015) 83–96
96