INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE

UNIVERSITY OF BRADFORD
Faculty of Engineering & Informatics
School of Engineering
GEOTECHNICAL ENGINEERING: INGENUITY IN SOIL
REINFORCEMENT METHODS AGAINST SCOUR FAILURE
Edison Derrick Mugoya
12006149
BEng (Honours) Civil & Structural Engineering
Project Supervisor: Dr MHA Mohamed
Stage 3 BEng Final Year Project
April 2015

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Edison Derrick Mugoya Final Year Project
INGENUITY IN SOIL REINFORCEMENT METHODS
AGAINST SCOUR FAILURE

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PROJECT AIM
To study the behaviour of various soils to enhance the stability and strength of abutments using soil
reinforcement methodologies.
PROJECT OBJECTIVES
To encourage the use of discrete natural fibres methods to enhance the slope stability on
embankments.
To investigate the use of Geosynthetics materials to increase the cohesion between soil types as an
attempt to increase the overall strength of the soil.
To use soil reinforcement techniques to mitigate scour failure around bridge abutments.
To exploit the flow around an abutment to understand different scour conditions.

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DECLARATION OF INDEPENDENT WORK
Student Name: EDISON DERRICK MUGOYA
UoB Number: 12006149
Course: BEng (Honours) Civil & Structural Engineering
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Table of Contents
INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE ...................i
PROJECT AIM.......................................................................................................................................ii
PROJECT OBJECTIVES .......................................................................................................................ii
DECLARATION OF INDEPENDENT WORK ...................................................................................iii
Acknowledgements................................................................................................................................ix
Chapter 1 Introduction..........................................................................................................................10
Chapter 2 Literature Review.................................................................................................................11
Soil Characteristics and Properties ...................................................................................................11
Soil Classification.........................................................................................................................11
Sedimentation ...............................................................................................................................12
The determination of soil moisture ...............................................................................................15
A New View of Abutment Scour......................................................................................................17
Stress Conditions for Failure ............................................................................................................17
Effective and neutral stresses........................................................................................................21
Scour conditions................................................................................................................................25
Shear Strength...................................................................................................................................29
Stability of Earth Slopes ...................................................................................................................35
Clay Slopes ...................................................................................................................................35
Stability analysis of the infinite slope...........................................................................................37
Consolidation Theory........................................................................................................................45
Soil Compaction................................................................................................................................46
Factors Affecting the Compaction Process...................................................................................47
Different Compaction Methods ....................................................................................................47
Chapter 3 Case Study............................................................................................................................49
Case study: New Zealand bridge scour experiences.........................................................................49
Bulls Road Bridge.........................................................................................................................49
Qualitative Analysis of Expected Scour Development.................................................................51
Definitions.....................................................................................................................................52
Motivation for review ...................................................................................................................52
Abutment form and construction ..................................................................................................56
Abutment form..............................................................................................................................56
Abutment layout............................................................................................................................57
Abutment construction..................................................................................................................57

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Pier proximity ...............................................................................................................................59
Sediment and soil boundary material............................................................................................60
Flow field......................................................................................................................................61
Design scour depths ......................................................................................................................64
Estimation of scour depths............................................................................................................64
An Essential design question ........................................................................................................65
Influence of pier proximity ...............................................................................................................67
Other scour processes....................................................................................................................67
Chapter 4 Methodology ........................................................................................................................68
Embankments Design and Application.............................................................................................68
Soil Reinforcement Techniques........................................................................................................69
Soil Reinforcement by fibre materials ..........................................................................................69
The main factors that make Natural Geotextiles useful................................................................69
Jute Fibres.....................................................................................................................................70
Coir Geotextiles ............................................................................................................................71
Bamboo and Timber Fibres...........................................................................................................71
Combination of Geotextiles ..........................................................................................................72
Soil Reinforcement by Geosynthetics...........................................................................................72
Chapter 5 Implementation of Soil Reinforcement ................................................................................73
DIRECT SHEAR TEST ...................................................................................................................73
What is this test about...................................................................................................................73
Purpose:.........................................................................................................................................74
Standard Reference:......................................................................................................................74
Apparatus:.....................................................................................................................................74
Equipment:........................................................................................................................................75
Test Procedure ..............................................................................................................................77
Chapter 6 Results & Analysis...............................................................................................................78
Chapter 7 Conclusions and Recommendations.....................................................................................85
References.............................................................................................................................................86
Books ................................................................................................................................................86
Journals.............................................................................................................................................86
Websites............................................................................................................................................87
Appendix...............................................................................................................................................89
Appendix A: Project Management and Organisation .......................................................................89
Appendix B: Plan of Action..............................................................................................................90
Appendix C: Gantts Chart.................................................................................................................91

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Appendix D: Mind Map....................................................................................................................93
Appendix E: ......................................................................................................................................94

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Table 1: Generally accepted soil classification (Online, Reference) ....................................................12
Table 2: Calculation of the coefficient of uniformity ...........................................................................14
Table 3: Fellenius’s Construction for centre of Rotation......................................................................36
Table 4: Maximum Dilation angle for all densities, stress and fibre conditions...................................83
Table 5: Angle of friction and Cohesion intercept for all series of tests...............................................83
Figure 1: Soil Classification Chart (Online, Reference www.nrcs.usda.gov).......................................13
Figure 2: Grain-size curves (Online, Reference www.fao.org) ............................................................14
Figure 3: The Neutron Moisture Meter; modern and conceptual images (Online, Reference
www.usyd.edu.au) ................................................................................................................................16
Figure 4: Apparatus used to demonstrate difference between effective and neutral stress
(http://www.iitbhu.ac.in/ internet reference).........................................................................................21
Figure 5: Effective stress when there is no water flow (http://www.iitbhu.ac.in/ internet reference) ..24
Figure 6: A butment-scour conditions: Scour Condition A - hydraulic scour of the main channel bed
causes bank failure, which causes a failure of the face of the abutment embankment (a); Scour
Condition B - hydraulic scour of the floodplain causes failure of the ..................................................26
Figure 7: Field example of Scour Condition A.....................................................................................28
Figure 8: Field example of Scour Condition B.....................................................................................28
Figure 9: Field example of Scour Condition C for a wing-wall abutment............................................29
Figure 10a: Stress conditions at failure.................................................................................................30
Figure 11: Stress conditions including envelope ..................................................................................30
Figure 12: Particle in contact causing locking......................................................................................31
Figure 13: example of shear failure in soils..........................................................................................31
Figure 14: Shear Box apparatus for test on shear strength....................................................................32
Figure 15: Triaxial test apparatus for test on shear strength .................................................................33
Figure 16: Stability of clay slope..........................................................................................................35
Figure 17: Method of Slices for clay Slopes.........................................................................................37
Figure 18: Failure condition for an infinite slope of cohesionless soil .................................................38
Figure 19: Failure condition of an infinite slope of cohesive soil.........................................................40
Figure 20: Slip circle: Cohesive soil.....................................................................................................41
Figure 21...............................................................................................................................................42
Figure 22: Friction circle method..........................................................................................................44
Figure 23: Optimum Moisture Content graph.......................................................................................47
Figure 24: Schematic of long, multi-span bridge over a compound channel........................................53
Figure 25: Schematic of relatively short bridge over a narrow main channel ......................................53
Figure 26: Abutment scour resulting in embankment failure by collapse due to geotechnical
instability. .............................................................................................................................................54
Figure 27: Scour at I-70 bridge over Missouri River from 1993 flood. Flow was from left to right.
(Photo from Parola et al. 1998).............................................................................................................55
Figure 28: Plan views of the two common abutment forms: (a) Wing-wall; (b) Spill-through (Ettema
et al. 2010). ...........................................................................................................................................56
Figure 29: Definitions of embankment length, floodplain width, and main channel width (Ettema et al.
2010).....................................................................................................................................................57
Figure 30: an isometric view of the geometry used for spill-through abutments. ................................58
Figure 31: The geometry and dimensions of a standard-stub abutment commonly used for spill-
through abutments (prototype scale indicated); design provided by the Iowa DOT (Ettema et al. 2010)
..............................................................................................................................................................58

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Figure 32: The geometry and dimensions of a wing-wall abutment - compacted earth fill embankment
extends back from the abutment structure (prototype scale indicated); design provided by the Iowa
DOT (Ettema et al. 2010)......................................................................................................................59
Figure 33: depicts an example of a bridge with a pier located close to an abutment............................60
Figure 34: Variation of soil and sediment types at a bridge crossing (Ettema et al. 2010)...................60
Figure 35: Flow structure including macro-turbulence generated by flow around abutments in a
narrow main channel. (Ettema et al. 2010). ..........................................................................................61
Figure 36: Flow structure including macro-turbulence generated by floodplain/main channel flow
interaction, flow separation around abutment, and wake region on the floodplain of a compound
channel. (Ettema et al. 2010). ...............................................................................................................62
Figure 37: For a spill-through abutment well set back on a flood-plain, deepest scour usually occurs
where flow is most contracted through the bridge waterway. ..............................................................63
Figure 38: Interaction of flow features causing scour and erodibility of boundary (Ettema et al. 2010).
..............................................................................................................................................................63
Figure 39:: A common situation of abutment failure; scour has led to failure and partial washout of the
earth fill spill-slope at this abutment. A basic question arises as to how abutment design should take
scour into account. ................................................................................................................................66
Figure 40: Failure of abutment fill in September 2009 Georgia flood accompanied by failure of
approach roadway (Hong and Sturm 2010). .........................................................................................66
Figure 41: Typical Earth Dike with Drain ............................................................................................68

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Acknowledgements
The subject matter of this project appealed to me as I was studying geotechnical engineering
in the University of Bradford 2012-2015 and in response I favoured the idea of reading articles and
journals on the related topic of soil mechanics as an effort to enhance the structures for a sustainability
feasibility study on Whitby’s upper harbour endeavour wharf centre. I thought that I could spot the
house that had sunk slightly due to the geology and not as a cause of seismic activity. Many sections
of this project consists of the accumulations of study’s done in the field of soil reinforcement and
I.C.E and I.Struct have been of major assistance in the up to date goings of the new findings in
methods to improve the technique of strengthening soil mass. The author to whom I acknowledge
most of the study from is Karl Terzachi from his book theoretical soil mechanics tenth print 1943 and
well referenced to Harvard University in appreciation of its liberal encouragement of the pursuit of
knowledge, this book was gratefully dedicated in assembling the acquired understanding of
Geosynthetics toward soil reinforcements.
Journals such as shear behaviour of a geogrids-reinforced coarse-grained soil based on
Reynolds, h.r. and protopapadakis, p heavily assisted me in the make-up of my literature review. Also
the laboratory evaluation of governing mechanism of frictionally resistance was assisted by Craig, r.f
soil mechanics based of the degree of problems in soil mechanics and foundation engineering
problems picked up from Menzies and Simon’s journal. The core sections of experimental and
numerical analysis of large scale pull out tests also found in chapter 4 methodology was
acknowledged mainly by babu sivakumar, g.l. study. The implementation section was an extract from
the journal on soil bioengineering/biotechnical stabilization of slope failure written by engineer, haley
& Aldrich, Inc., Cambridge, Massachusetts and was acknowledged by principal, robbin b. sotir &
associates Inc., Marietta, Georgia.
I would also like to acknowledge my supervisor Dr. M.H.A. Mohamed for his continuous
efforts in guiding my written work evidence of our meetings can be found in the logbook attached at
the appendix shows the proof of the meetings and minutes. Being the head of discipline at the faculty
of engineering & informatics at the university of Bradford he has advised me and guided me through
the structure of the project.

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Chapter 1 Introduction
The concept of soil reinforcement is an ancient technique and is demonstrated abundantly in
nature by animals, birds and tree roots. It is an attempt to improve the stability of the soil such as the use
of plants and their roots to simply hold the ground together to prevent the top soil from sliding.
Constructions using these techniques are known to have existed in the fourth and fifth millennia B.C.
(Jones, 1985). This concept has been favourable for many centuries and has been used for improvement
of certain desired properties of soil such as bearing capacity (𝑞 𝑢), shear strength (φ) and permeability to
name a few. The idea in principle was first developed by Vidal (1969), by inducing an element into soils
as a means of increasing its shear resistance to lateral loads. According to Vidal’s idea the interaction
between soils and the reinforcing specimen was to increase its horizontal resistance due to friction
generated by gravity, in this case the unit weight of soil. The first retaining wall built in France 1968 was
based off Vidal’s idea. The technology became wide spread through to USA and Europe but wasn’t
picked upon by Asian countries such as India due to the high cost and availability of the reinforcing
materials. More recently, the soil reinforcement techniques is well established and is used in a number of
applications such as to control drainage, improve soil bearing capacity and prolonging effects of scour
failure.
For a vast majority of reinforced soil structures the main reinforcement force is axial tension.
Earth Structures by themselves are weak, apply a tensile force on the earth structure and it will fail with
minimal effort. The horizontal forces acting on a soil mass can provide enough force to overcome the
resultant forces over the obliquity for the given soil type. However inserting tensile reinforcements in the
horizontal plain will enable vertical faced masses of soil to remain stable. (Jones, 1985) stated that the
mode of action of reinforcement in soils is not through carrying developed tensile stresses, as in
reinforced concrete, but rather of anisotropic reduction of the normal strain rate. A wide range of
materials since (Jones, 1985) have been used and tested as reinforcing materials. Early structures used to
be formed using organic materials such as timber, straw, reed, bamboo and sisal for reinforcements also
known today as fibre materials. Although they are generally considered as less durable they meet the
functionality of its design and are cheap. New materials such as steel, concrete, glass fibre, rubber,
aluminium, and thermo-plastics have been used more successfully due to their material properties.
One of my objectives to this study is to review the present state of knowledge regarding bridge-
abutment scour and evaluate the leading methods currently used for estimating design scour depth. This
study should ease my dissertation in formulating an improved approach to the fight against scour failure.
It focuses on research information obtained since 1990, and that must be considered in updating the
scour-depth estimation methods recommended by AASHTO1, and used generally by engineering
practitioners. Reinforcing soils provides a relatively cheap form of construction for retaining walls,
bridges abutments, marine structures, reinforced slopes and embankments mainly due to the ease and
speed of construction. These methods are also considered to be more aesthetically pleasing to the eye as
they retain the natural beauty of the landscapes whistle using up minimal land.
The main aim is to study the behaviour of various soils such as clay and sands to find new ideas
and methods to enhance the overall slope stability around embankments and to strengthen abutments
using soil reinforcement methodologies. In this study I will also exploit the shear strength, tensile strength
and bearing capacity behaviour of various soils to draw to a conclusion on a more versatile, reliable, cost
effective method to reduce geotechnical related accidents such as landslides that are a result of
overloading or poor engineering techniques.

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Chapter 2 Literature Review
This section covers the background knowledge that is necessary to draw any solution to scour
failure by looking at the fundamentals of soil mechanics. The complexity of bridge abutment scour
necessitates a thorough evaluation of the physical processes involved and their parameterization in scour
depth estimation formulas. As river flow approaches a bridge, the streamlines converge due to the
physical contraction in width and then diverge once through it. Understanding the soil properties and
characteristics prior to construction of bridge abutment could prevent scour failure
Soil Characteristics and Properties
Generally Speaking classifying soil characteristics and its properties in the field of geotechnical
engineering has not been adequately appreciated. Henry R. Reynolds and P. Protopapadakis mentions that
is essential to become familiar with the results of soil tests and to grasp the significance of its properties
and behaviours through experience. Soil moisture content, plastic limit and liquid limit are just some of
the basic fundamentals towards soil reinforcement. Soil reinforcement I believe is a study that will
enhance and change modern living styles. As the push for sustainability is crucial this increases the
demand for structures that are able to be stable and constructed in a manner that is suitable to the
environment. Engineering practices are faced with challenges ; whether a certain soil is suitable for
tipping in an embankment of a specified height, whether the load of the structure is capable of being
supported without undue or even settlement, whether groundwater lowering or artificial cementations for
deep excavation is possible, and other similar foundation problems.
The next few sections is allocated to examples involving soil classification and soil properties
which will in turn provide the basis for the practical engineering problems contained in the later chapters.
Soil Classification
Soil consists of a mixture of mineral practical’s and water, and includes a wide range of materials
from shingle to plastic clay. Soil is any uncemented or weakly cemented accumulation of minerals
particles formed by the weathering of rocks. Most soil can be easily excavated by hand or hand tools. In
the study of soil mechanics it is most important to be able to classify the different soils into defined types
based on their size, shape and nature of the particles. However it also must be recognized that the nature
of the soil particles is largely dependent on its moisture content. The classification tests are two types:-
(1) Mechanical analysis, by means of sieving or sedimentation to determine the size-distribution
of the soil particles.
(2) Index tests, for the soils passing a 36-mesh B.S. sieve, by means of which the type of soil is
deduced from moisture content at standard consistencies.
The generally accepted term and standards are given in the table 1 and table 2 together with the
chart in figure 1.

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Table 1: Generally accepted soil classification (Online, Reference)
Sedimentation
The grains of soil settle in a liquid with a velocity which may be calculated by Stoke’s Law,
which states that the rate at which small sphere sinks in liquid is directly proportion to the square of the
diameter of the sphere (Reynolds & Protopapadakis, 1959). This law applies only when considering grain
diameters between 0.2mm and 0.0002mm. Grains larger than 0.2mm diameter settle with a varying
velocity and particles less than 0.0002mm diameter are in colloidal suspension. Velocity of settling in
cms.per sec.
Equation 1: Stoke’s Law
𝑣 =
2(𝜌 − 𝜌 𝑤)
9𝑛
(
𝑑
2
)
2
Where 𝜌 denotes specific gravity of the soil grains,
𝜌 𝑤 denotes specific gravity of the liquid,
N denotes coefficient of viscosity for the liquid, (0.000103
kg.sec.per.sq.m for water at 20°)
D denotes the diameter of the soil grains

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Figure 1: Soil Classification Chart (Online, Reference www.nrcs.usda.gov)
The simplified formula for the spherical particles descending in still water is as follows:-
Equation 2: Stoke's Law for Spherical Particles
𝑣 = 8800𝑑² 𝑜𝑟 𝑑 =
√ 𝑣
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The above equations can used to solve the following problems. Please refer to the appendix for the
solutions to the problems below.

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1. How long would it take for a particle of soil 0.01mm in diameter to settle from the surface to the
bottom of a pond 10ft. if the specific gravity of the water is 1.0 and of the soil is 2.55, and the
coefficient of viscosity of water is 0.1025g.sec.per sq.m? (ans:0.0088cm.per sec using the above
equation)
Figure 2: Grain-size curves (Online, Reference www.fao.org)
Figure 2 above shows the grain-size vs percentage of fineness by weight. The coefficient of uniformity is
the ratio of the particle size for 60% finer by weight to the effective diameter. From inspection of figure 2
the following table can be drawn up:-
Table 2: Calculation of the coefficient of uniformity
Sample Effective diameter for
10% finer
Diameter for 60%
finer
Coefficient of
uniformity
A 0.007mm 0.055mm 0.055
0.007
= 7.86
B 0.096mm 0.146mm 0.00.146
0.096
= 1.52
C 0.056mm 0.63mm 0.63
0.056
= 11.25
Soil Properties; Voids ratio, porosity, moisture content and density.
A Soil is made up of soil particles with voids between the particles filled with either moisture or
air, or both. The Natural moisture content of soil is determined by weighing a sample before and after
drying at 105°C. The loss in weight is expressed as a percentage of the dry weight. When soil is saturated,
the moisture content (𝑤𝑐), voids ratio (e), porosity (n).
To determine the specific gravity of soil particles, a pycnometer bottle is used. This consists of
flask which has a volume of 500c.c. at a certain temperature, usually 20°C, and this volume is marked on

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the neck of the bottle. A soil sample of 25 to 50 grs. Is placed in the bottle, which is then filled up with
distilled water. The liquid is boiled to expel the air the air adhering to the soil particles, and when cool the
bottle is filled up to the mark on the neck and weighed.
The specific gravity of the soil grains,
Equation 3: Specific Gravity of Soil
=
𝑤𝑒𝑖𝑔ℎ𝑡 𝑜𝑓 𝑠𝑜𝑖𝑙 𝑖𝑛 𝑡ℎ𝑒 𝑏𝑜𝑡𝑡𝑙𝑒
{𝑤𝑔𝑡. 𝑜𝑓 𝑑𝑟𝑦 𝑠𝑜𝑖𝑙 𝑖𝑛 𝑡ℎ𝑒 𝑏𝑜𝑡𝑡𝑙𝑒} + {𝑤𝑔𝑡. 𝑜𝑓 𝑏𝑜𝑡𝑡𝑙𝑒 + 𝑤𝑎𝑡𝑒𝑟} + {𝑤𝑔𝑡. 𝑜𝑓 𝑏𝑜𝑡𝑡𝑙𝑒 + 𝑠𝑎𝑚𝑝𝑙𝑒 + 𝑤𝑎𝑡𝑒𝑟}
The determination of soil moisture
(Reynolds & Protopapadakis, 1959) had research by Saskatchewan research council and prairie
Road builders section of Canadian construction association based from the determination of soil moisture
and density by nuclear radiation that gave rise to the neutron moisture meter used today to determine soil
moisture on site. Paper was presented to the American society of testing materials in Cleveland, Ohio by
Messrs. D. A. Lane, B.B. Torchinsky and J.W.T. Spinks on the development of the technology.
The moisture meter is based on the fast neutron bombardment of soil, which results in the
reflection of slow neutrons and their action on a detector of indium foil. The apparatus consists of a cone-
shaped head which is lowered into a 2-in. diameter vertical tube sunk into soil. Attached to this head is a
capped aluminium cylinder about 11’’ In length. A neutron source is placed in the position by means of
electro-magnets within the cylinder and holder contain indium foil is lowered into its seating by means of
cord, thus positioning the indium foil around the neutron source. (Figure 3).
The fast neutrons emitted by the neutron source lose their energy by collision with the hydrogen
atom in the water molecules in the soil and therefore, become slow neutrons which are absorbed by the
stable indium foil causing a change to radioactive indium. The slow neutrons are more readily absorbed
than fast neutrons and the activity induced in the foil during a given period depends upon the activity
induced in the foil during a given period now depends upon the amount of moisture in the soil. An
equation has been evolved which relates the amount of activity of the soil to the soil moisture content.
The foil exposed for a definite period, withdrawn in its holder and placed round the Geiger tube of a
portable meter, and a reading is taken at the termination of the further definite period of time. However in
practise a specific procedure of a ten-minute exposure with a further one-minute delay before
measurement has been chosen as a standard routine. An initial reading of the foil should be taken before it
is used and additional five or six identical foils are required for the use, so that a reasonable period of time
can elapse before their re-use. Generally speaking for practical purposes, the zone of influence can be
considered to be a sphere of six inch radius.
The determination of soil density is based on the amount of gamma-ray absorption. The density
meter relies on the amount of gamma-ray absorption of soil which is proportional to its density. The same
neutron source is utilised as before in conjunction with the Geiger tube of the portable meter, but the
cylinder is larger, being about 20ins. Long with a lead shield place between it and the neutron source as
protection to direct gamma radiation. Thus the only rays which can strike the tube are those through the
soil. The tube is connected to the portable rate meter which indicates the average rate at which gamma
rays reach the tube and this value can be calibrated against wet density. It is essential, however that the
relative positions of the source and Geiger tube are constant throughout the test in order to produce
accurate and comparable calibration curves. A number of field experiments have been carried out using

16
aluminium tubes and steel pipes, and it is only necessary to adjust the calibration curves according to the
material used.
These results were checked with oven-drying methods and all determinations for moisture content agreed
within at least 3%, although most of the results were within 2%. Tests from the ground that is flooded
gave similarly effective results, but where soils have organic composition some variation does occur. The
presence of organic material affects the meter in the same way as if water were present and a correction
factor must be introduced dependent upon the soil type and further experimental work is taking place in
this field. Another possible use for neutron meter is to determine the asphaltic content of bituminous
pavement composition. The activation of the meter is similar to that for determining the water content and
compensating calibration curve can be developed to indicate the asphaltic percentage present. The use of
neutron and density meters necessitates certain precautions being taken by the operator. An electromagnet
mounted on a 3-ft. length of
5
16
𝑖𝑛𝑐ℎ diameter rod projecting from an ordinary flashlight case facilitates
the safe handling of the 75Ib. When the neutron source is placed in position the operator must not stand
directly over it and must carry a film monitor indicating the exposure to gamma rays, so that he may keep
within the allowable margin.
Figure 3: The Neutron Moisture Meter; modern and conceptual images (Online, Reference www.usyd.edu.au)

17
A New View of Abutment Scour
Based on the foregoing definition of scour and documentation in this report of numerous failures
of bridges due to abutment scour, one of the important initial findings is that many abutment failures
occur due to scour and sliding of the earth fill embankment on the main stream side of the abutment into
the scour hole, or outflanking due to erosion of the earth fill embankment on the floodplain side due to
overtopping or inadequate drainage protection. Even more difficult to evaluate is the vulnerability to
scour caused by lateral shifting of the channel thalweg such that it directs flow adversely towards
abutments and embankments. Whereas much of the laboratory research of recent years has focused on
solid abutments that extend into the soil foundation, such as with sheet piles or other fairly rigid
foundations, more attention should be focused in the future on erodible embankments. Recognition of the
difference between erodible and solid abutments provides a factor for classifying existing scour prediction
formulas and introduces the importance of geotechnical failure caused by hydraulic scour. In addition, it
suggests the need for estimating the strength of the embankment over the range of construction forms
varying from unprotected, compacted soils of various types through rock riprap revetment to the solid
abutment, and incorporating this estimate into a more comprehensive scour prediction formula. These
considerations pose a fundamental design problem in that partial failure of the embankment that occurs as
sliding of earth fill and/or riprap into the scour hole may ultimately reduce the total scour depth while
complete failure of the embankment may be intolerable if it results in failure of the bridge approach slab
or the first bridge span. This leads on to the importance of understanding the stress condition for failure of
various soil types.
Stress Conditions for Failure
Relation between normal stress and shearing resistance makes up the better part of this section.
Stress will be referred to as force per unit of area of the section of soil mass. It has been genuine to
assume that the relation between normal stress (𝜎) on every section through a mass of cohesive soil and
the corresponding shearing resistance (s) per unit area can be represented by an empirical equation.
Equation 4: Coulomb's empirical equation
𝑠 = 𝑐 + tan 𝜙
Provided (𝜎) is compressive stress. The symbol (c) represents the cohesion, which equals to the
bearing shearing resistance per unit area if ( 𝜎) = 0. The equation is known as coulomb’s equation. For
cohesion-less soils (c = 0) the corresponding equation is
Equation 5: Coulomb's equation
𝑠 = 𝜎 tan 𝜙
The Values c and 𝜙 contained in the preceding equations can be determined by means of
laboratory test, by measuring the shearing resistance on plane sections through the soil at different values
of the normal stress 𝜎. In practise we are quite interested in the shearing resistance of saturated or almost
saturated soils. A change of stress in a saturated soil is always associated with some change of water
content. The rate of the change of water content produced by a given change of state of stress depends on
several factors, including the degree of permeability of the soil. If the stresses which ultimately lead to
failure of the test specimen are applied more rapidly than the corresponding change in the water content
of the specimen can occur then parts of the applied normal stress 𝜎 will be carried at the instant of failure,
by excess hydrostatic pressure which is required to maintain the flow of the excess water out of the voids

18
of the soil. At a given value of 𝜎 the part of 𝜎 which is carried by both the value c and φ depend not only
on the nature of soil and its initial state but also on the rate of stress application, on the permeability of the
material, and on the size of the specimen. The value of 𝜙 obtained from such tests is called the angle of
shearing resistance. For clays this angle can have any value up to 20° (exceptionally more) and for loose,
saturated sands any value up to 35°. In other words, no definite value can be assigned to the angle φ for
any soil, because it depends on conditions other than the nature and initial state of soil.
On the other hand if the stresses on the test specimen are applied slowly enough, the normal stress
𝜎 which acts on the surface of the slider at the instant of failure is almost entirely transmitted from grain
to grain. Tests of this kind are known as slow shear tests. The rate at which such tests must be made
depends on the permeability of the soil. If shear tests on sand with a given initial density are made in such
a manner that the stresses are entirely transmitted from grain to grain, we find that the shearing resistance
s = 𝜎 tan 𝜙 is practically independent of the character of the changes of stress which preceded the failure.
For instance, it makes practically no difference whether we increase the unit load on the sample
continuously from 0-1 ton per square foot and then reduce it to 1 ton per square foot. If the load on the
sample at the instant of failure is equal to 1 ton per square foot, the resistance s is the same in both cases.
In other words, the shearing resistance s depends solely on the normal stress on the potential surface of
sliding. A shearing resistance of this type is called a frictional resistance and the corresponding value of
φ represents an angle of internal friction. Within the range of pressure involved in engineering problems
the angle of internal friction of sand can usually be considered constant for practical purposes. Its value
depends on the nature and initial density of the sand. It varies between the extreme limits of about 30° and
50°. The difference between the angle of internal friction of given sand in the densest and in loosest state
may be as high as 15°.
In (Terzaghi, 1962) investigations of soil problems he generally assumed that the angle of internal
friction of sand is identical with the angle of repose described in Article 3. However, as stated above,
laboratory experiments have shown that the angle of internal friction of sand depends on a large extent on
the initial density. In contrast to the angle of internal friction, the angle of repose of dry sand has fairly
constant values. It is always approximately equal to the angle of internal friction of the sand in the loosest
state. BS EN and text books also contain a list of values for the angle of repose of cohesive soils, although
as shown in Article 4, the angle of repose such soils depends on the height of the slope.
When equation 5 is used in connection with stability computations the value 𝜙 always represents
the angle of internal friction of sand. This can only be expressed with the sufficient accuracy by equation
4,
𝑠 = 𝑐 + 𝜎 𝑡𝑎𝑛𝜙
In order to find out whether the term 𝜎tan𝜙 satisfies the requirements for a frictional resistance
i.e. whether the resistance 𝜎tan𝜙 depends solely on the normal stress𝜎, we submit our material with a
given initial water content to two different tests.
In on test we increase 𝜎 from zero to 𝜎1 and determine the corresponding shearing resistance𝑠1.
In the second test, we consolidate our materials under a pressure of 𝜎2 which is greatly larger than 𝜎1 and
finally we determine by means of slow shear test, the corresponding shearing resistance 𝑠′1. The process
of temporarily keeping a sample under pressure which is higher than ultimate pressure is known as
preconsolidation. Experiment show that the shearing resistance 𝑠′1 of preconsolidated material may be
equal to if not greater than𝑠1. If the two values are equal 𝜎 𝑡𝑎𝑛𝜙 in equation 1 represents a frictional
resistance and we are justified considering 𝜙 an angle of internal friction. On the other hand if 𝑠′1 is

19
greater than 𝑠1, we know that the resistance 𝜎 𝑡𝑎𝑛𝜙 represents the sum of a frictional resistance and some
other resistance which is independent𝜎. The most conspicuous permanent change produced by
preconsolidation consists in an increase of the density of the material and a corresponding reduction of
water content. If 𝑠′1 is appreciably greater than 𝑠1 we always find the water content corresponding to 𝑠′1
is lower than that corresponding to 𝑠1. We know from experience that the value c in the equation 1
increases for a given clay with decreasing initial water content. Therefore in most cases we are justified in
drawing the following conclusion. If 𝑠′1 is appreciably greater than 𝑠1, the resistance 𝜎 𝑡𝑎𝑛𝜙 in the first
part is friction produeced by normal stress 𝜎 and the second part is the increase of the cohesion the
reduction of the water content which occurred while the pressure on the specimen was increased from
zero to 𝜎. (Terzaghi,1962) statement can be expressed by the equation below.
Equation 6: Terzaghi 1962
𝑠 = 𝑐 + 𝜎 𝑡𝑎𝑛𝜙 = 𝑐 +
𝜎1 + 𝜎3
2
𝑁 + 𝜎𝑡𝑎𝑛𝜙 𝑓
Whereby 𝜎1 and 𝜎3 represent the extreme principal stresses at failure after the slow test and N is
an empirical factor. The fraction 𝜎𝑡𝑎𝑛𝜙 𝑓 of the shearing resistance changes with orientation of a section
through a given point, while the fractions c and
𝜎1+𝜎3
2
𝑁 are independent of the orientation. The customary
methods for experimentally investigating the shearing resistance of cohesion soils merely furnish the
values c and 𝜙 on the left-hand side of the equation. The determination of the values 𝜙 𝑓 and N requires
elaborate supplementary investigations which belong in the realm of soil physics.
For cemented sand the value 𝑠′1 is usually very close to that of 𝑠1. For such materials the value
𝜎 𝑡𝑎𝑛𝜙 in equation 1 represents only a frictional resistance. On the other hand when experimenting with
clay we find that the shearing resistance 𝑠′1 of the preconsolidated sample is always appreciably greater
than 𝑠1 at the same load. Hence in connection with clay the angle of 𝜙 in equation 1 represents neither an
angle of internal friction nor a constant for the clay, even when its value has been determined by means of
slow shearing tests. If one makes a series of slow tests on clay with a given initial water content after
increasing the pressure on the samples from zero to different values𝜎1, 𝜎2 etc. one gets an equation
If one makes another series of tests on specimens of the same material after preceding
consolidation of the samples under a pressure which is higher than the test pressures one gets another
equation
𝑠 = 𝑐′ + 𝜎 𝑡𝑎𝑛𝜙′
Whereby c’ is considerably higher than c and 𝜙′ is considerably smaller than 𝜙. Hence when
using coulomb’s equation 1 in connection with clays, note that c and 𝜙 contained represent merely two
empirical coefficients in the equation of a straight line. The term cohesion is retained only for historical
reasons. It is used as an abbreviation of the term apparent cohesion. In contrast to the apparent cohesion,
the true cohesion represents the part of shearing resistance of soil which is a function only of the water
content. It includes not only c in coulomb’s equation but also an appreciable part of 𝜎 𝑡𝑎𝑛𝜙. There is no
relation between apparent and true cohesion other than the name.

20
To help you visualize the difference between apparent and real cohesion I’ll need to consider a
material that has increases it’s cohesion with compaction. Clay is a good generic example. By making a
series of shear tests with the material we obtain the following:-
However when we look at the parts of shearing resistance of the material that is due to cohesion
we obtain the equation below also derived from Mohr circle,
𝑠 = 𝑐 +
𝜎1 + 𝜎3
2
𝑁 + 𝜎 𝑡𝑎𝑛𝜙 𝑓
If you compared the two preceding equations we get that the true cohesion of the material is equal
to c but to:

21
𝑐 𝑒 = 𝑐 +
𝜎1 + 𝜎3
2
𝑁
If the entire pressure on the clay is transmitted from grain to grain the true cohesion is never
smaller than the apparent cohesion. Therefore if 𝜎 𝑡𝑎𝑛𝜙 from the equation 1 is equal to zero then
𝑠 = 𝑐
For liquids the values c and 𝜙 are zero which means that
𝑠 = 0
Effective and neutral stresses
In real life the voids of every fine-grained soil are partly or wholly filled with water. If we take a
section through a saturated soil, then parts of it will pass through the solid particles and part of it through
the water. In order to assure the mechanical implications of this fact, we need to consider the test
arrangement as shown below in figure 4. This apparatus represents a section through a layer of cohesion
less soil which occupies the bottom of a vessel. At the outset of the test the free water level is supposed to
be located immediately above the surface of the soil and the layer is assumed to be
So thin that we neglect the stress due to the
weight of the soil and water which are loacted
above the horizontal section known as the datum
the (ab). If we raise the water level to an elevation
of ℎ 𝑤 above it’s orginal place the normal stress
expereince on the section (ab) increses from
almost zero to 𝜎 = ℎ 𝑤 𝛾 𝑤 where 𝛾 𝑤 is the unit
weight of water. Yet this may increse the
compressicve stress from practically zero to 𝜎 on
every horizontal section. In which the soil doesn’t
make a measurable compression of the layer of
soil. On the other hand if we increse the intenstiy
of the pressure on the layer by the same amount
the 𝜎, by loading the surface o fthe layer with
lead shot the resulting compression of the layer is
very appreciable. As seen practised in the field of
the Shanghai lotus river to date. This appartus can
also demonstrate that the postio of water level in
the vessel has no influence on the shearing
resistance s of the soil, whereas an equivalent
soild surchage increases the shearing resistance
a b
ℎ 𝑤
Figure 4: Apparatus used to demonstrate
difference between effective and neutral stress
(http://www.iitbhu.ac.in/ internet reference)

22
very much. These and many experiments similar to it lead to the conclusion that the compressive stress in
saturated soil consists of two parts with very different mechanical effects. One of which equals to the
pressure in the water produces neither a measurable compression nor a measurable increase of the
shearing resistance. (Terzachi, 1962) called it neutral stress 𝑢 𝑤 . He quotes that the product of the unit
weight of water 𝛾 𝑤 and the height ℎ 𝑤 to which the water rises in a piezometric tube at the point under
consideration, then the corresponding equation is
𝑢 𝑤 = ℎ 𝑤 𝛾 𝑤
The height as shown in figure4 represents the piezometric head at the point of observation. It can
be positive or negative. Hence how the neutral stress can be a positive or negative value. If it is positive it
is usually called the pore-water pressure.
Subsurface water is divided into zones of positive and negative pore pressures. The dividing line
is the groundwater table (also known as phreatic surface) where the pressure is equal to atmospheric
pressure. Below the groundwater table, the soil is fully saturated, and the pore pressure is above
atmospheric pressure and positive in value. Above the groundwater table where the soil is unsaturated, the
pore pressure is below atmospheric pressure and hence is negative in value. In this zone, the pore water is
continuous or semi continuous and the pore water pressure is below atmospheric pressure. The magnitude
of the negative pore pressure (sometimes called soil suction) is controlled by surface tension at the air-
water boundaries within the pores and is governed by grain size. In general, the finer the soil particles, the
larger the saturation capillary head, and hence the higher the negative pore pressure. Rainfall infiltration
from the ground surface may rapidly reduce the magnitude of negative pore pressure. Any change in these
pore pressures alters alter the shear strength of soil and therefore has a tremendous effect on the slope
stability.
The water level measured in a piezometer within the saturation zone coincides with the water
table. However, the pore pressures are no longer hydrostatic if there is a flow. In this instance, the pore
pressure from any point within the soil mass is computed by means of a flow net, from the difference in
head between the point and the free water surface.
By lowering effective stress, positive pore pressure reduces the available shear strength within the
soil mass thereby decreasing the slope stability. Increase in positive pore pressure can be rapid after a
period of heavy rainfall. That is a major reason why many slope failures occur after heavy rainfall. The
rate of increase, however, depends on many factors such as the rate of rainfall, the nature of the ground
surface, the catchment area, and the soil permeability. Pore pressure below the groundwater table can be
assessed using analytical, numerical, and graphical methods. Various analytical methods are available for
determining flow nets and pore pressure distributions in slopes. Numerical techniques using finite
difference or finite element method provide powerful tool for obtaining pore water distributions in slopes.
They are the only means by which transient flow situations can be fully modelled.
Negative pore pressures increase the effective stresses within a soil mass and improve the
stability of a slope. (Ho and Fredlund, 1982) suggested increase in shear strength due to negative pore
pressure as

23
Equation 7: Ho and Fredlund, 1982
𝑐 = 𝑐′ + (𝑢 𝑎 − 𝑢 𝑤 )𝑡𝑎𝑛𝜙 𝑏
Where c = total cohesion of the soil
c’= effective cohesion
(𝑢 𝑎 − 𝑢 𝑤 ) = matrix suction
𝜙 𝑏 = the slope of the plot of matrix suction when - is held constant
Here, a matrix suction (𝑢 𝑎 − 𝑢 𝑤 ) increases the shear strength by (𝑢 𝑎 − 𝑢 𝑤 )𝑡𝑎𝑛𝜙 𝑏. The
increase in soil strength can be represented by a three-dimensional failure surface using stress
variable (𝜎 − 𝑢 𝑎 𝑎𝑛𝑑 𝑢 𝑎 − 𝑢 𝑤 ), as shown in Figure. These negative pore pressures reduce in magnitude
when the degree of saturation increases and become zero when the soils are fully saturated; the major
problem in evaluation of stability in unsaturated soils is associated with the assessment of reduction in
negative pore pressure and possible increase in positive pore pressure as a function of rainfall history.
Let us consider full saturation of the rock, including the joint, where no drainage of water is
allowed. If we assume that water is incompressible and that no flow of water into or out of the joint is
allowed, the volume of the test specimen including the joint must remain constant. Under this condition,
the water must sustain stresses sufficient to prevent volume change of the specimen. The total applied
stress across the joint will be transmitted by the rock asperities and by the water. If the water carries some
of the normal stress, then the rock asperities carry less normal load and therefore has less shear strength
than it would be if drained. The normal stress transmitted by the water is equal to the joint water pressure.
The stress transmitted through the rock asperities is, therefore, equal to the applied stress minus the joint
water pressure. The joint shear strength will now be reduced proportionally. The reduced normal stress
acting through the rock contacts is termed as the effective normal stress and is given by
𝜎′ 𝑛 = 𝜎 𝑛 − 𝑢
Where by
𝜎′ 𝑛 =effective normal stress
𝜎 𝑛 = normal stress
𝑢 =water pressure
The total stress imposed on such a soil will be sustained by the soil; the effective stress, and the
pore pressure, u. A reservoir can be used to create an upward seepage through the soil sample. For this
purpose, we assume that the valve leading to the upper reservoir is closed. Thus, there is no water flowing
through the soil sample (figure 4). This is the case of no seepage the Effective stress is
𝜎′ 𝑛 = 𝐻2(𝛾𝑠𝑎𝑡 − 𝛾 𝑤)
Upward seepage conditions can be induced in the laboratory using constant-head permeability
test apparatus, in figure 5. The upper reservoir causes the water to flow upward through the soil sample. If
the hydraulic gradient is large, the upward-seepage force will cause the effective stress within the soil to

24
become zero, thus causing a sudden loss of soil strength in accordance with the effective-stress principle.
However, if downward seepage is allowed, effective stress sigma' is
𝜎′ 𝑛 = 𝐻2(𝛾𝑠𝑎𝑡 − 𝛾 𝑤)
Figure 5: Effective stress when there is no water flow (http://www.iitbhu.ac.in/ internet reference)
In the analysis of stability of slopes in terms of effective stresses, the pore water pressure
distribution is of fundamental importance and its evaluation is one of the prime objectives in the early
stages of any stability study.

25
Scour conditions
One method for classifying abutment scour depends on abutment location in a channel, the
relative erodibilities of sediments forming the main-channel bed and soils forming the floodplain (see
Figure 45), as well as to the shear strength of the compacted earth fill forming the approach embankment.
In addition, other conditions such as stream morphologic changes and lack of control of highway runoff
can lead to abutment scour under unexpected and less well-defined circumstances.
Three common conditions of abutment scour
Figure 49-c illustrates the three scour conditions for spill-through abutments:
1. Scour Condition A. Scour of the main-channel bed, when the channel bed is far more erodible than the
floodplain. Figure 49a illustrates how scour of the main-channel bed causes the main-channel bank to
become geotechnically unstable and collapse. The collapsing bank undercuts the abutment and
embankment, which in turn collapses locally. Soil, and possibly riprap, from the collapsed bank and
embankment slide into the scour hole;
2. Scour Condition B. Scour of the floodplain around the abutment. This condition also is equivalent to
scour at an abutment placed in a rectangular channel, if the abutment is set back from the main channel.
As the amount of bed-sediment transport on a floodplain usually is quite low, this scour condition usually
occurs as clear-water scour. Figure 49b shows that the floodplain scours around the abutment, and
especially slightly downstream of it. The scour hole locally destabilizes the embankment side slope,
causing embankment soil, and possibly riprap, to slide into the scour hole; and,
3. Scour Condition C. Scour Conditions A and B may eventually cause the approach embankment to
breach near the abutment, thereby fully exposing the abutment column. For this condition, scour at the
exposed stub column essentially progresses as if the abutment column were a pier, as illustrated in Figure
49c. For the same reasons as given for Condition B, this scour condition usually occurs as clear-water
scour.

26
Figure 6: A butment-scour conditions: Scour Condition A - hydraulic scour of the main channel bed causes bank failure, which
causes a failure of the face of the abutment embankment (a); Scour Condition B - hydraulic scour of the floodplain causes failure
of the

27
The three scour conditions may occur also for wing-wall abutments. However, a couple of additional
erosion processes can result in failure of the main-channel bank and the approach embankment:
1. The local flow field generated at the corners of the abutment can cause local scour at those locations;
and,
2. Exposure of the piles beneath the abutment pile cap can cause river-bank and embankment soil to be
eroded out from beneath the pile cap.
Provided no substantial geotechnical failure of the abutment occurs for scour Conditions A and B,
scour deepens to an equilibrium level commensurate with the abutment flow field’s capacity to attain a
balance with the rate of sediment inflow to the scour region (live-bed scour) or the channel boundary’s
resistance to erosion (clear-water scour).
A scour event (or series of events) at an abutment, may involve a sequence of all three scour
conditions, resulting in several local maxima for scour depth for a wing-wall abutment. When an
abutment is close to the main channel, Condition A may develop relatively quickly, with Condition B
occurring at a slower rate. Either, or together, Scour Conditions A and B may eventually cause the
approach embankment to undergo a slope-stability failure. If the embankment extensively washes out, so
as to expose the abutment structure, scour may then develop at the abutment structure as if the abutment
were a form of pier (Condition C). Accordingly, an important design consideration is that the stub or
wing-wall abutment should not fail when exposed; i.e. foundations of wing-walls should be deep enough
that the wing-walls do not fail when exposed to a pier-like scour condition.
For design estimation of scour depth, it is useful to consider the likely rates or sequences in which
the three scour conditions developed, and to ask -- What is the greatest scour depth that reasonably could
occur near the abutment? Will that scour depth pose a slope-stability problem for the earth fill
embankment adjoining an abutment foundation or for the floodplain bank of the main channel? What is
the deepest scour that could occur at the abutment column foundation itself, and does that scour occur
when the embankment is breached so as to fully expose the abutment column? The set of photographs in
Figures 50 through 51 depict situations where Scour Conditions A, B, and C occurred at bridge
abutments.

28
Figure 7: Field example of Scour Condition A
Figure 8: Field example of Scour Condition B.

29
Figure 9: Field example of Scour Condition C for a wing-wall abutment.
Shear Strength
Soil failure may occur as a result of exceeding the maximum shear stress that the soil can
handle/sustain. This is a major factor in understanding soil behaviour in evaluating its influence on soil
reinforcements. When it comes to the analysis of the stability of soil masses, whether at a point or on any
plane within a soil mass the shear stress becomes equal to the shear strength of the soil then failure will
occur at this point. Prior to the principles of effective stress, the shear strength 𝜏 𝑓of soil at a point on a
particular plane was expressed by coulomb as a linear function of the normal stress at failure 𝜎𝑓 on the
plane at the same point.
Equation 8: Shear Strength
𝜏 𝑓 = 𝑐 + 𝜎𝑓 𝑡𝑎𝑛𝜙
Where c and 𝜙 are shear strength parameters referred to as the cohesion intercept and the angle of
shearing resistance, respectively. However, in accordance with the principle that governs shear stress with
in soil it can be resisted only by the skeleton of solid particles, shear strength should be expressed as a
function of effective normal stress a failure 𝜎′ 𝑓 the shear strength parameters being giving as c’ and 𝜙′.
𝜏 𝑓 = 𝑐′ + 𝜎′
𝑓 𝑡𝑎𝑛𝜙′
Failure will thus occur at any point in the soil where a critical combination of shear stress and
effective normal stress develops. It should be expected that c’ and 𝜙′ are simply mathematical constants
defining a linear relationship between shear strength and effective normal stress. It easier to think of these
stress as inter-particle forces; therefore, if effective normal stress is zero then shearing resistance must be
zero (unless there is cementation between the particles) and the value of c’ would be zero. This brings me
to this point of crucial to the implementation of shear strength parameters.
In most 2 dimensional representations we plot shear stress 𝜏 against effective normal stress 𝜎′.
This is a state that can be represented either by a point with coordinates𝜏 𝑎𝑛𝑑 𝜎′, or by a Mohr circle
defined by the effective principles stresses with coordinates𝜎′1 𝑎𝑛𝑑 𝜎′3. As shown below the stress
points and Mohr circles represents stress states at failure (fig6a & fig7).

30
Figure 10a: Stress conditions at failure.
The line through the stress points that are touching the Mohr circle may be straight or slightly
curved and is referred to as the failure envelop. A state that means any stress points that are plotted above
this line is impossible as the soil would be in motion; either a twist or turning one. There two methods to
quantify the shearing strength parameters. (1) The envelope is represented by the straight line as defined
in the equation above. From which the parameters c’ and 𝜙′ can be obtained. These are referred to as
tangent parameters and are only valid over a limited range. It has been used for since Coulombs
discovered it. If the failure envelope is slightly curved the parameters are obtained from a straight line
ShearStress(MPa)
Normal Stress (MPa)
Figure 11: Stress conditions including envelope

31
approximation to the curve over the stress range. It should be noted that the use of tangent parameters
does not reflect the real life scenario.
From this point I will be covering the shear strength of sand and the shear strength of clay as
found from reconstituted specimens in the laboratory. When sand particle arrangement is confined
laterally, strains can only in the vertical direction, such as in an oedometer test. As the vertical stress is
increased small groups of particles in the loose sand will collapse to the surrounding voids producing a
volumetric (or vertical) stain on the soil. This produces a more tightly packed arrangement with which a
larger number of particles are now in contact with each other. This makes it more fixed in place as it is
given less freedom to move about. This is a phenomenon known as locking. Thus the vertical stress-strain
curves are concaved. The term stiffness represents the gradient of this curve so it can be seen that the soil
is becoming stiffer. As we increase the vertical stress that is in contact between the particles, then the
particles will begin to facture and crush producing yield and allowing vertical strains to increase. An
increase in the number of particles (fig 8) due to crushing will produce a further particle contact which in
turn reduce the average contact stress between particles and causes the stiffness to continue increasing
due to further locking.
Figure 12: Particle in contact causing locking
The level at which structural collapse, locking and yielding occur depends on the initial overall
density and inherent strength or for a lack of a better word crushability of particles. Under confined
compression where horizontal strains are restricted, as the vertical stress increases the horizontal stress
increases. The horizontal stress does not need to be as large as the vertical stress this is so; because parts
of the latter will be supported by the shearing resistance of the sand. (Jaky, 1944) claimed that the
horizontal stress to vertical ratio denoted as 𝑘0 at rest, or lateral strain condition is as follows: 𝑘0 = 1 −
𝑠𝑖𝑛𝜙. Whereby 𝑘0 is the coefficient of earth’s pressure at rest.
Figure 13: example of shear failure in soils

32
The failure as shown in figure 9 above is due to inadequate strength at shear interfaces. Soil
derives most of its strengths from cohesion and frictional resistance. Cohesion is a measure of the forces
that cement the particles together. The above images could well be dry sand with no cementation, dry
sand with some cementation, soft clay or stiff clay. As mentioned above Coulombs law based on the
Mohr’s Failure criteria it suggests that the figure 9 suffered from a lack of good adhesives beneath the
road that would give an increased strength and protection against slides. Vertical drains are not visible
here so I think the water couldn’t escape easily loosening the soil by introducing a pressure difference
from the side to the other making an active pressure scenario 𝜎ℎ < 𝜎𝑣 . Such case means that the soil is
pushing outwards downslope, causing a rotational slip as seen in figure 9.
Shear box test and triaxial test are good measures of the shear strengths of soils. When a direct
measure of the shear strength of a granular soil is desired a shear box test is often used. Although the
results are from reconstituted data with densities and particle arrangements that are different from those
found in situ an allowance should be made for this. Nowadays the shear box test tends to be used for
investigation of shear strength properties of the more unusual granular materials where correlation
between 𝜙 and in situ tests such as SPT or CPT are not available or unreliable. These include crushable
sands like calcareous, vesicular sands, granular fills (fragmented rock particles both soft and hard, waste
materials (colliery spoil) and the shear strength of interfaces between two construction materials (steel
and sand, steel piles) or plastic and clay (geomembrane and clay liner). Other applications which have
utilised this test are for quick undrained strength of clay and cut-plane or reversal test (returning the split
specimen to its starting point) for the determination of the drained residual strength.
Figure 14: Shear Box apparatus for test on shear strength

33
Figure 15: Triaxial test apparatus for test on shear strength
The triaxial test apparatus was first developed in the 1930s and has larger been replaced by the
direct shear test in commercial laboratories. It consists of applying a shearing stress with a cylindrical
sample of soil by changing the principles stresses𝜎1 𝑎𝑛𝑑 𝜎3. The commonest procedure is to keep the
triaxial cell pressure 𝜎3 constant (𝜎2 = 𝜎3 with axial symmetry) and increasing the axial or vertical stress
𝜎1 until failure is achieved. The essential feature as shown in figure 11 below is standardized at 38mm
and 100mm diameter specimens. It has a height: diameter ratio of 2:1 to ensure that the middle section of
the specimen is free to shear. If this ratio is less than 2:1 then shear stresses at the end of the sample in
contact with the platens will affect the results by constraining the failure planes. The soil specimen is
surrounded by rubber membrane to stop the cell fluid from entering the soil and altering its moisture
content. For weaker soil specimens a correction to account for the restraint provided by the membrane
should be applied. The axial stress is applied by a motorised drive which raises the specimen and the cell
against the piston reacting on a load frame. A proving ring or load ring or load cell between the piston and
load frame measures the axial force F from which the principle stress difference or deviator stress,
𝜎1 − 𝜎3, is calculated using the equation below.
𝐹
𝐴
= 𝜎1 − 𝜎3
The strength of the soil is obtained from Mohr circle plot. AS the stress is applied the specimen often
becomes barrel-shaped so the vertical stress in the middle of the specimen must be determined from the
force measured and this increased area by applying an area correction to each reading (Barnes, 2010).
The corrected area A in the middle of the specimen is obtained for a drained test from:
𝐴 = 𝐴0
(1−𝜀 𝑣)
(1−𝜀 𝑎)
Where 𝐴0 is the initial cross-sectional area, 𝜀 𝑎 is the vertical or axial strain and 𝜀 𝑣 is the volumetric strain.
For an undrained test where the volumetric strain is zero (𝜀 𝑣=0) the corrected area is given by

34
𝐴 =
𝐴0
(1−𝜀 𝑎)
Now tying in to what I’ve mentioned above I conclude this section with the residual strength in
soil. Although the critical state strength is often referred to as the ultimate strength this condition is
achieved with homogenous shearing, i.e. all the samples are undergoing the same shear strain and these
strains are not excessively large. It has been seen particularly from the studies of the old landslips done by
(Skempton, 1964) where significant straining has occurred on thin shear surfaces that the operative shear
strength on these surfaces was much lower than the critical state strength. What I mean is the residual 𝜙 𝑟′
value for London Clay for example can be as low as 10° whereas at the critical state 𝜙𝑐𝑣′ is greater than
20°. (Barnes, 2010) it’s essential therefore to identify the presence or otherwise, pre-existing slip surfaces
in clay soil on a sloping site. Small changes have been noted in the surface topography and also in the
pore pressure conditions from earlier sections of this project. Residual strength is then attained when large
shear strains have occurred on the thin zone or plane of sliding in clay soil. This formation is where the
clay particles have been arranged to produce a firm preferred orientation in the direction of the slip
surface. (Lupini, 1981) recognised three modes of residual shear behaviour;
 Turbulent
This occurs where behaviour is dominated by rotund particles. For soils dominated by platy
particles with high inter-particle friction this mode may also occur. In this mode energy is
dissipated by particle rolling and translation. No preferred particle orientation occurs and residual
strength still remains high so that 𝜙 𝑟′ can be taken as𝜙𝑐𝑣′.
 Sliding
When behaviour is dominated by platy, low friction particles, sliding occurs on a shear surface
with strongly oriented particles and the strength is low. 𝜙 𝑟′ depends mainly on the mineralogy
coeffeicent of inter-particle friction 𝜇 and pore water chemistry.
 Transitional
This involves turbulent and sliding behaviour in different parts of a shear zone. The residual shear
strength can be obtained using a ring shear apparatus as (Bishop, 1971 and Bromhead, 1978)
found. A ring-shaped thin sample of remoulded soil is sheared in a direct shear manner by
rotating the upper half of the sample above the lower half with sufficient strain until a slip surface
is formed on which the lower strain is measured from the torsion applied. As illustrated in figure
6 the residual strength 𝜏 𝑟 is related to the normal stress 𝜎 𝑁
′
applied on the slip surface by:
𝜏 𝑟 = 𝜎 𝑁
′
𝑡𝑎𝑛𝜙 𝑟′
Although for many soils the plot of 𝜏 𝑟 vs 𝜎 𝑁
′
shows a small cohesion intercept 𝑐 𝑟′ or a curvature
of the plot will show me that the stress range is applicable to the site conditions and therefore must be
determined by 𝜙 𝑟′. In majority of the cases, if the clayey content is 40-50% or more or the plasticity index
is 30-40% or more then the 𝜙 𝑟′ value can be expected to be lower than 15° (Lupini et al, 1981).

35
Stability of Earth Slopes
In the design scope for sandy soils, the angle I made by the slope with the horizontal should be
smaller than the angle of internal friction of sand, 𝜙. Normally in loose sands the angle of friction is about
32°, but this angle increase to 40° with very dense sands. It is important to remember that the angle of
slope for stability of cohesionless soil is independent of the height, which may be indefinite. Furthermore,
the weight of the material doesn’t affect the stability of the slope; therefore the safe angle for a submerged
sand slope is the same as that for one composed of dry sand, with the exception of the special case of
damp sand, which has a high angle of repose due to capillary attraction. Special conditions exist with
partially submerged sand slopes affected by tidal conditions which may cause the stability of the fine sand
slope to be considerably less than that of dry sand. Assuming the angle of safe slope with the horizontal is
I then for submerged cohesionless soil slope
𝑇𝑎𝑛 𝑖 =
𝜌 − 1
𝜌 + 𝑒
𝑡𝑎𝑛𝜙
Conditions which exist with submerged slopes subject to sudden draw-downs; may be caused in a similar
way with embankments of fine sand exposed to rainstorms sufficiently heavy to result in saturation of the
sand fill.
Clay Slopes
A slip which has taken place in clay slope has three definite characteristics for example a crack
appears at the top of the bank, a portion of the material in the bank slips downward and there is a heave at
the toe as shown below.
Figure 16: Stability of clay slope
In a bank of homogeneous clay materials the slip line failure in the slope closely follows the arc of circle,
and for stability
𝑊 × 𝑑 = 𝐿 × 𝑠 × 𝑅
Where W = weight of the segment of soil of unit thickness, L = length of arc segment, R = radius of the
cylindrical surface of shear, d = distance of the centre of gravity of the segment from a vertical through

36
the centre of curvature and s = average intensity of shear resistance per unit area of the cylindrical
surface. From the above equation it can be observed that in order to investigate the stability of clay slope
it is necessary to ascertain the weight of the soil, the apparent cohesion and the angle of internal friction
of soil concerned. These data may be obtained from shear tests as shown in earlier sections under the
heading shear strength.
It must be noted and emphasised that whistle the height of sand slopes are entirely different and
safe slope is a function of height. Sands possess an angle of repose; whistle clays do not have such a
characteristic, although their behaviour is measured by their shear strength. If for intense the factor of a
clay slope is F, then
𝐹 =
𝐿 × 𝑠 × 𝑅
𝑊 × 𝑑
The factor of safety cannot be considered as something absolute, as slope with factors of safety
less than unity have proved to be stable, but in the design of new works for cuttings and embankments it
is advisable to maintain a safety factor between 1.25 and 1.5. The centre of critical circle is found by trial
and error for the minimum value of the factor of safety, and the following notes will assist in the
determination of centre for this circle. So if the shear strength increases with depth then the slopes are
steeper than 45°, then we can use the following table3 for values of angle 𝛼 𝑎𝑛𝑑𝛽 to find the centre of the
arc in figure 12.
Table 3: Fellenius’s Construction for centre of Rotation
Slope Angle of Slope (°) Angle 𝜶 (°) Angle 𝜷 (°)
1-0.58 60 29 40
1-1 45 28 37
1-1.5 33-47 26 35
1-2 26-34 25 35
1-3 18-26 25 35
1-5 11-19 25 37

37
If slopes are flatter than 45°, or if the clay is homogenous, then the centre of the critical circle lies on
vertical through the mid-point of the slope. The circle tends to be deep and would tangent at an
underlying layer of harder clay is such a stratum existed. However when making adjustments for the
centre of the critical circle, horizontal movements are more likely sensitive than vertical one.
Figure 17: Method of Slices for clay Slopes
Figure 13 shows a problem involving a railway cutting that was made in clay which tended to
increase in shear strength with depth. The soil has a weight 120 lb.per sq. cu.ft and an average shear
resistance of 600 lb.per.sq.ft. It was assumed that the cutting was 6m, and it is proposed to adopt slope of
1 to 1.5. Using the values from table 3 we can ascertain the centre of the critical circle and calculate the
safety factor for the proposed slopes. O the centre of circle, to the arc from the toe of the slope to the top
of the vertical section 8. The weight of each slice is proportional to the centre ordinate and the average
height of each slice as plotted vertically below the curvature.
Stability analysis of the infinite slope
The limit equilibrium method is used for the analysis of finite slopes. Slopes extending to infinity
do not exist in nature. For all practical purposes any slope of great extent with soil conditions essentially
same for all identical depth below the surface are known as infinite slopes. Infinite slopes in dry
sand: The figure 14 shows the failure conditions for an infinite slope of cohesionless soil.
The factor of safety of an infinite slope is defined as the ratio of soil strength in the required shear stress
for equilibrium. The factor of safety against sliding is given by

38
𝐹 =
𝜏 𝑓
𝜏
=
tan(𝜙)
tan(𝑖)
Where 𝜏 𝑓=shear strength
𝜏 = mobilized shear strength due to gravity
𝜙 = angle of internal friction
i = inclined angle of slope.
Figure 18: Failure condition for an infinite slope of cohesionless soil
Infinite slope in 𝑐 − 𝜙 soils in 𝑐 − 𝜙 soil, the slope is stable as long as the slope angle i is equal
to or less than the angle of internal friction 𝜙. If the slope angle i, is greater than 𝜙, the slope can be stable
only upto limited height known as critical height is given by
𝐻𝑐 =
𝐶
𝛾(tan(𝑖) − tan(𝜙))𝑐𝑜𝑠²𝑖
Where C= cohesion
𝛾= unit weight

39
𝜙= angle of internal friction
i = slope angle
𝐻𝑐= critical height
If the factor of safety Fc is applied in cohesion, the mobilized cohesion at depth H, given by
Then the depth H calculated by using mobilized cohesion Cm will not be critical. The factor of safety
against height also represents the factor of safety with respect to cohesion Fc.
Fc is given by
A dimensionless parameter called a stability number is often useful for the analysis of slopes of 𝑐 −
𝜙 soils and can be defined by the following equation
Where 𝑠 𝑛 = stability number (a dimensionless quantity). The reciprocal of stability number is known as
stability factor.

40
Figure 19: Failure condition of an infinite slope of cohesive soil
Stability analysis of finite slopes; Failure of finite slopes occurs along a curved surface. In
stability analysis of finite slopes, the real surface of rupture is replaced by an arc of a circle. As to the
mode of failure, the slope may fail basically in the following two ways; 1 the failure surface passing
through the toe of the slope or above the toe of slope is known as slope failure. 2 the rupture is deep
seated and passes through the embankment supporting soil below the toe of the slope is known as base
failure.
1. Slope failure above toe (or face failure)
2. Slope failure through toe (or toe failure)

41
3. Base failure
The base failure generally occurs particularly when the soil beneath the embankment is softer and
more plastic than the slope forming soil itself. There are several methods available for the stability
analysis but the following methods are simple and widely practiced.
1. Slip circle method (Swedish circle method)
2. Friction circle method
Swedish circle method or method of slices: The method assumes the surface of sliding is an arc of
a circle. This was established by studying the failure or embankments in Sweden. (fig.16)
Analysis of the purely cohesive soils ( = 0 analysis)
Consider a likely circular slip surface AD (fig.16) with centre at O.
Figure 20: Slip circle: Cohesive soil

42
The disturbing moment of the cylinder of the soil about O= Wx. Where ‘x’ is the distance of the
line of action of W from the vertical line passing through the centre of rotation. If Cu is the unit cohesion,
L= length of the slip arc,
𝐴𝐷 =
2𝜋𝛾𝛿
360
The shear resistance developed along the slip surface will be equal to Cu.L. The resisting moment
preventing the soil from moving is all due to friction along arc length AD which has a lever arm equal to
radius r about O. Resisting moment = cohesion x arc length AD x r = CuL.r Factor of safety against
sliding = (Resisting moment / Disturbing moment)
=
𝑐 𝑢 𝐿𝑟
𝑊𝑥
A series of slip circles are checked, and the lowest factor of safety is the likely failure plane.
𝑐 − 𝜙 Soil (𝑐 − 𝜙 Analysis). In order to study stability of the slope of a 𝑐 − 𝜙 soil, a possible slip circle is
chosen and divided into strips of equal width as shown in fig 17 (a) and (b).
a) Slip circle: Friction soil
Figure 21
b) One strip

43
Consider one strip as shown in figure 17(b). The forces between the slices are neglected. Vertical
weight W can be considered in two components (1) at right angles to arc of circle (normal component)
and (2) tangential to arc of circle (tangential component).
Disturbing moment about centre O = T x r
Total driving moment = 𝑟Σ𝑇
Where Σ𝑇 = algebraic sum of all tangential components
The resisting force on one strip is made up of cohesion and friction and is given by:
Resisting force= 𝑐Δ𝐿 + 𝑁𝑡𝑎𝑛 (𝜙)
Resisting moment on one strip= 𝑐Δ𝐿 + 𝑁𝑡𝑎𝑛 (𝜙)
Total resisting moment = [𝑐ΣΔ𝐿 + 𝑡𝑎𝑛 (𝜙)Σ𝑁]𝑟
[𝑐ΣΔ𝐿 + 𝑡𝑎𝑛 (𝜙)Σ𝑁]𝑟
Where Σ𝑁 = sum of all normal components, L= arc length
Factor of safety
𝐶𝐼 + tan(𝜙)Σ𝑁
Σ𝑇
Friction circle method (Fig 17. (a) and (b)): The friction circle method of stability analysis of
slope is applicable to 𝑐 − 𝜙 soils. The friction circle method also assumes the failure surface as the arc of
a circle.

44
Figure 22: Friction circle method
Fig.18 shows a failure arc of radius r with O as the centre. In the friction circle method of analysis
of 𝑐 − 𝜙 soil, the resultant reaction vector R at an obliquity of 𝜙 to an element of the failure arc will be
tangential to the small circle of radius kr.sin 𝜙. The small circle of radius kr.sin 𝜙 is therefore called the
friction circle.
In the 𝜙 circle system with a known 𝜙, the following quantities are known.
i. The magnitude and direction of weight of sliding wedge (W)
ii. Direction of resultant reaction (R)
iii. The direction of total cohesion CI (parallel to the chord, I = AC)
To determine the magnitude of R and cohesion Cm, force triangle is constructed in which
magnitude of W is known. The factor of safety with respect to cohesion based on the assumption that
frictional strength has been fully mobilized, is given by
𝐹
𝑐=
𝐶
𝑐 𝑚
A number of slip circles are analysed and the lowest factor of safety is the likely failure plane.
Stability of the slopes of earth dam: Earth dams must be safe against slope and foundation failure
for all operating conditions. There are three generally recognised critical stages based on pore pressure fir
which the stability of the embankment should be ascertained. These three situations are (i) end of
construction, (ii) steady-state seepage and (iii) rapid drawdown.
Usually construction pore pressure reaches their maximum values when the embankment reaches
maximum height. After the reservoir has been filled for a long time, pore pressure is determined by steady
state seepage conditions and may be estimated by the construction of flow net. Rapid lowering of the
reservoir produces the third critical situation, particularly for low permeable soils. The upstream slope
stability can be critical for the construction of rapid drawdown condition. The downstream slope should
be checked for the construction and steady state seepage condition.

45
Consolidation Theory
Previously the effect of seepage was assumed that the volume occupied by the water per unit of
volume of the soil was independent of the state of stress in the soil. If this condition was satisfied the
quantity of water which flows out of an element of soil such as a, flow represented in hydrostatic pressure
conditions on four sides of a prismatic element of sand, is equal to the quantity of water which enters the
element, regardless of whether or not the state of stress in the soil changes. This condition, known as
continuity condition, is expressed in a mathematical terms by the differential equation as shown below.
Equation 9: Differential equation
𝜕𝑣 𝑥
𝜕𝑥
𝜕𝑥𝜕𝑧𝜕𝑦 +
𝜕𝑣𝑧
𝜕𝑧
𝜕𝑧𝜕𝑥𝜕𝑦 = 0
There is no real soil which strictly satisfies the continuity condition, because every change in the
state of stress produces a certain change in the volume of voids,Δ𝑛, per unit volume of soil. Yet if the soil
is very permeable and not very compressible, the change of the porosity due to a change in the state of
stress in the soil can usually be disregarded.
(Terzachi, 1943) stated that a change effective stresses in a highly compressible soil, such as clay
or sand-mica mixture, is likely to produce an important change Δ𝑛 in the volume of voids 𝑛. Hence if the
voids of such a soil are completely filled with water and remain in that state a change in effective stresses
involves a change in water content of soil. Every process involving a decrease of the water content of a
saturated soil without replacement of water by air is called a process of consolidation. The opposite
process is called a process of swelling, which involves an increase of water content due to an increase of
the volume of voids. A further complication arises if the soil combines high compressibility with low
permeability. Both of these properties are exhibited to a high degree by fat clays. In soils with such
characteristics, changes to the water content due to a change in the state of stress takes place very slowly,
because the low permeability of soil does not permit a rapid transfer of the water from one part of the
mass of soil to another or to an adjoining highly permeable, compressible stratum. This phenomenon
produces a time lag between a change of the external forces which acts on a feebly permeable,
compressible stratum and the corresponding change of the water content of soil. It is the principal cause
of the progressive settlement of foundation on clay and of many other processes of outstanding practical
importance.
Some assumptions where made in the theory of consolidation with a few exceptions to all
existing theories of consolidation prior to 1943. The assumptions being that the voids of the soil are
completely filled with water; both the water and solid constituents of the soil are perfectly
incompressible; Darcy’s law is strictly laid; the coefficient of permeability k is a constant; and the time
lag of consolidation is due entirely too low permeability of soil. The theories contained in the following
articles are used based on the following supplementary assumptions, unless a departure from these
assumptions is specifically mentioned. The clay is laterally confined; both the total and the effective
normal stresses are the same for every point of any horizontal section through the clay and for every stage
of the process of consolidation; an increase in the effective pressure from an initial value 𝑝̅ 𝑜 to a final
value 𝑝̅ reduces the void ratio of the clay from an initial value 𝑒 𝑜 to a final value e; the ratio
𝑎 𝑣𝑐 =
𝑒 𝑜 − 𝑒
𝑝̅ − 𝑝̅ 𝑜
𝑔𝑚−1
𝑐𝑚²

46
Is assumed to be a constant for the range of pressure 𝑝̅ 𝑜 to 𝑝̅. It is called the coefficient of
compressibility. If the effective pressure is reduced from an initial value 𝑝̅ to a final 𝑝̅’ the void ratio
increases from an initial e to a final e’. The ratio
𝑎 𝑣𝑐 =
𝑒′ − 𝑒
𝑝̅ − 𝑝′̅
𝑔𝑚−1
𝑐𝑚²
Is also assumed to be a constant for the range of pressure 𝑝̅ 𝑜 to 𝑝̅’. It is called the coefficient of
elastic recovery. From the first equation we obtain the following
𝑎 𝑣𝑐(𝑝̅ − 𝑝̅ 𝑜) = 𝑒 𝑜 − 𝑒
The quantity 𝑒 𝑜 − 𝑒 represents the decrease of volume of the voids in a block of soil with the
initial volume 1+𝑒 𝑜. The initial voids ratio 𝑒 𝑜 corresponds to a volume of voids per unit volume of soil
𝑛 𝑜 =
𝑒 𝑜
1+𝑒 𝑜
and the final voids e to 𝑛01 =
𝑒
1+𝑒 𝑜
therefore the decrease ∆𝑛 of the volume of voids per unit of
the initial volume of the soil is denoted as
∆𝑛 = 𝑛0 − 𝑛01 =
𝑎 𝑣𝑐
1 + 𝑒0
(𝑝̅ − 𝑝̅0) = 𝑚 𝑣𝑐(𝑝̅ − 𝑝̅0) = 𝑚 𝑣𝑐∆𝑝̅
Wherein ∆𝑝̅ is the increase of the effective unit pressure. The value
𝑚 𝑣𝑐 =
𝑎 𝑣𝑐
1 + 𝑒0
𝑔𝑚−1
𝑐𝑚²
This is called the coefficient of volume decrease. The corresponding value for a process of
swelling due to a reduction of the effective pressure is
𝑚 𝑣𝑠 =
𝑎 𝑣𝑠
1 + 𝑒0
This is called the coefficient of volume expansion. If there is no possibility of misunderstanding
the second subscript on the symbol 𝑎 𝑣 and 𝑚 𝑣 will be omitted. The preceding assumptions determine the
physical properties ascribed to the ideal clay which constitutes the subject of the subsequent theoretical
investigations. The equation listed above represents a crude approximation of the relation between the
effective pressure on real clay in a state of complete lateral confinement and the corresponding void ratio.
Soil Compaction
Soil compaction means increasing soil density that makes working with soil easy, helps in
erecting stable structures, and reduces maintenance costs. Compaction of soil brings stability and strength
with it. Foundations fail most commonly because of improper compaction methods or poorly compacted
soil that allows water to seep through the foundation and cause structural damage. Implementing
mechanical methods to compact soil means densifying the soil, filling the pore spaces, improving the
shear resistance of soil, and providing better water movement through the soil particles. The compaction
process largely depends upon the type of soil you are dealing with because different soils have different
physical properties and accordingly different compaction methods should be adopted. Compaction also
prevents frost damage of soil and increases its durability.

INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE

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