1. Profits Incorporated
10488 Eastborne Avenue
Los Angeles, California, 90024
(510) 593­7394
March 13, 2015
Resort Development, Incorporated
2145 Mackenburger Drive
Los Angeles, California, 90040
Attn: Dr. Stewart, Dr. Hudson
Subject: Response to work authorization
Hotel development
South La Senda Drive and Bay Drive
Laguna Beach, California
Dear Doctor Stewart and Doctor Hudson,
In accordance with your authorization to proceed on February 3rd, 2015, this letter report
presents the results of our geotechnical investigation undertaken at the site of the proposed hotel
development, by Resort Development, Inc.
As per our understanding, the proposed hotel site, located in Laguna Beach, California, is
currently occupied by single­family homes and trailers.We understand the proposed construction
consists of a hotel with three portions: a central building with two wings; and a one­story
restaurant adjacent to the ocean bluff. All three portions of the hotel building are proposed to be
up to four stories above grade, with the basement of the main hotel approximately 25 feet below
the existing grade. The restaurant, however, is not planned to contain a basement. Your letter
listed the anticipated load associated with the central portion of the proposed hotel as 350 to 700
kips, with the southern wing as 200 to 350 kips, and with the western wing as 250 to 450 kips.
The proposed restaurant is expected to have relatively light column and wall loads. Additionally,
your letter contained a plan showing the layout and topography of the proposed hotel site, a mass
grading plan for the proposed hotel site and laboratory test results and boring logs from previous
investigations at the site by Moore and Taber in 1980, and AGRA Earth and Environmental, Inc.,
in 1999 (figures 1, 2, and appendices 1, 2).
1

2.
Figure 1:​ Site plan for proposed hotel location in Laguna Beach, Orange County, California,
showing locations of proposed hotel buildings, surface exposure of geologic units, locations of
geologic cross sections, and locations of borings.
2

3.
Figure 2:​ Mass grading plan for the site, showing location and and thickness of fill.
SCOPE OF WORK
The scope of work in this investigation included the following tasks:
1. A site investigation including the drilling of 7 borings: four using an 8” hollow stem
auger, and three using a 24” diameter bucket. Subsurface samples were taken during the
boring process using Shelby Tubes, Modified California Samplers (MCSs), and Standard
Penetration Tests (SPTs), and geologic cross sections were prepared at three locations at
the site.
2. Laboratory testing to characterize the engineering properties of both clayey and sandy
foundation soils, and of the compacted fill soil.
3

4. 3. Engineering analysis of clayey foundation soil settlement and bearing capacity under the
main hotel buildings; of bearing capacity of artificial fill under the hotel restaurant; of
slope stability of fill during placement; of lateral earth pressures on retaining walls and
basement walls (including consideration of bearing capacity, overturning, and sliding of
retaining walls); and of slope stability of the coastal bluffs.
4. Development of recommendations for slope stability during site excavation; for fill
placement, grading and compaction; for foundations under the main hotel buildings and
restaurant; for retaining and basement walls; and lastly for slope stability of the coastal
bluffs.
This report has been prepared for the exclusive use of Resort Development Inc. for
specific application to their proposed hotel site in Laguna Beach, Orange County, California in
accordance with generally accepted geotechnical engineering practices. No other warranty,
expressed or implied, is made. In the event that any changes in the nature, design, or location of
the facility are planned or made, the conclusions and recommendations contained in this report
should not be considered valid unless the changes are reviewed and conclusions of this report
modified or verified in writing.
The findings of this report are valid as of the present date. However, the passing of time
will likely change the conditions of the existing site due to natural processes or the works of
man. In addition, due to legislation or the broadening of knowledge, changes in applicable or
appropriate standards may occur. Accordingly, the findings of this report may be invalidated,
wholly or partly, by changes beyond our control. Therefore, this report should not be relied upon
after a period of three years without being reviewed by this office.
SITE CONDITIONS
Prior Investigations
The data from previous investigations at this site available to us at the time of our
investigation include twelve 6” flight auger borings from Moore and Taber performed in 1980
(B­5, B­6, B­7, B­8, B­12, B­13, B­15, B­17, B­20, B­24, B­27, B­28). and six 24” diameter
bucket borings from AGRA Earth and Environmental Inc. performed in 1999 (B­201, B­202,
B­203, B­204, B­206, B­210). The locations of these borings are shown in the plot plan (figure 1,
appendix 1), and the boring logs are attached as appendix 2.
Investigations Performed
Our site investigation included the drilling of 7 borings: four using an 8” hollow stem
auger, and three using a 24” diameter bucket (locations shown in the plot plan, appendix 1).
Subsurface samples were taken during the boring process using Shelby Tubes, MCSs, and SPTs.
The depths and sampling methods for samples taken from each boring are shown in the boring
logs (appendix 2). Samples taken during the SPTs were disturbed, and therefore used for grain
size analysis and Atterberg limit tests. All other laboratory tests were performed using the
relatively undisturbed samples from the Shelby Tubes and MCSs.
4

5.
Surface Conditions
The proposed hotel site is located along a coastal bluff in Laguna Beach, Orange County,
California, and is currently occupied by single family homes, trailers, and vegetation. The bluff
historically has stability problems resulting in semi­regular flow slides. Portions of the site are
paved with asphalt, and there is some surface exposure of local soil and rock, and pre­existing
fill. The current ground topography, the locations of the existing structures, vegetation, and
asphalt, and the surface exposure of various soil and rock units are shown in the existing plot
plan of the site (figure 1, appendix 1).
Subsurface Conditions
The boring logs performed as a part of this investigation were used in conjunction with
the boring logs from the investigations of More and Taber in 1980 and Agra Earth and
Environmental Inc. in 1999 to better constrain subsurface stratigraphy. Using the logs and the
geologic map, three geologic cross sections of the site (A­A’, B­B’ and C­C’) were prepared
(appendix 1). Section A­A’ and C­C’ strike roughly north, and section B­B’ strikes roughly
northwest.
The naturally­occurring lithology of the site is broken down into five units. Quaternary
beach sand (Qbs) and Quaternary landslide deposits (Qls) are the scarcest units, and are located
surficially along the coastal bluffs.
The next youngest unit in the stratigraphy section consists of Quaternary non­marine
terrace deposits (Qtn), primarily sandy silt, silty sand and silty clay. Upon inspection during
boring, the material was described as massive to poorly bedded, dark reddish brown in color,
porous, and well­graded with some subangular clasts.
Quaternary marine terrace deposits (Qtm) generally underlie the non­marine terrace
deposits, and consist primarily of clayey sand and silty sand. Upon inspection during boring, the
material was described as having poorly to well developed bedding, subangular to round clasts
up to 4” in size, and being light brown in color. The sand to silty sand in this layer has a
representative SPT blow count (N​1​)​60 ​of 18, a relative density (D​r​) of 63%, and an effective
stress friction angle (ϕ’) of 36.8° (table 3).
Lastly, the Tertiary San Onofre breccia (Tso) underlies the marine and non­marine
terrace deposits. The breccia was described during boring as being highly weathered and
oxidized, having massive to poor bedding, and being up to 60% angular clasts up to 3’ in size
mostly of schist.
Additionally, there is a pre­existing sandy fill throughout the site. However, there are no
records of placement for the fill. The majority of this fill, where it is identified in the plot plan
and borings, will be removed during the planned excavation and grading process. The exception
to this is the fill along the coastal bluffs, which will be excavated and refilled using properly
compacted fill.
Water seepage was encountered at a depth of 47’ in boring 3, and traces of water were
observed at a depth of 35’ in boring 6, 10 minutes after termination of drilling. Generally the
groundwater table (GWT) was found to be deep during drilling; however some analyses are
5

6. performed under the assumption of GWT rise during landscaping and irrigation of the hotel
grounds.
LABORATORY TESTING
Gradation Test
A bulk sample was collected from a test pit that cut through the existing fill, non­marine
terrace deposits, and marine terrace deposits at the borrow area. A water content test (ASTM D
2937) and sieve analysis (ASTM D 422) performed on material from this bulk sample yielded a
water content of 6.8%, and 32% of the sample passing the #200 sieve by weight, respectively
(figure 3). Sieve analysis was also performed on samples taken from boring three at 8 and 17 feet
depth (appendix 3).
Figure 3: ​Sieve analysis results for sample from borrow area on site.
Atterberg Limit Testing
The Atterberg limit test (ASTM D 4318), performed on a specimen from the bulk sample,
yielded a liquid limit of 19, a plastic limit of 14 and a plasticity index of 5 (appendix 3). The soil
was classified silty­clayey sand (SM­SC) using the Unified Soil Classification System (USCS).
Consolidation Tests
Consolidation tests (ASTM D2435­80) were performed on two samples of Quaternary
non­marine terrace deposits (Qtn) from boring 3, one taken at a depth of 8 feet, and one taken at
6

7. a depth of 17 feet (appendix 2). The test specimens were prepared by trimming a slice of the
material from the piston tube sampler down to 4­inch diameter by 7/8­inch tall cylinders. Figures
4 and 5 present the consolidation test results for the specimens, and table 1 summarizes the
resulting parameters.
Boring
No.
Sample
No.
Depth
(ft)
In­Situ
Moisture
Content
(%)
Dry Unit
Weight
(lbs/ft​3​
) e
σ'​v
(kips/ft​2​
)
σ'​P
(lbs/ft​3​
) γ​sat ​(lbs/ft​3​
) OCR
3 4 8 9.8 129 0.36 0.99 2.661
141.6
2.69
3 9 17 9.7 125 0.35 2.26 4.712 137.13 2.08
Table 1: ​Consolidation test results for two samples from boring 3.
Figure 4: ​Consolidation test results for a sample taken from boring 3 at a depth of 8 feet
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8.
Figure 5: ​Consolidation test results for a sample taken from boring 3 at a depth of 17 feet.
Shear Strength Triaxial Tests
A series of consolidated­undrained (CU) triaxial tests were performed on 5 specimens of
Quaternary non­marine terrace deposits prepared from the sample obtained from boring 3 at a
depth of 17 feet.
Specimen 1 Specimen 2 Specimen 3 Specimen 4 Specimen 5
σ'​c​ 1 (psf) 7000 8500 10000 7000 8200
σ'​c​ 2 (psf) 7000 8500 10000 2300 2700
Table 2: ​CU triaxial test isotropic consolidation stresses during consolidation (σ'​c​1) and during
shearing (σ'​c​2).
All specimens were allowed to consolidate under an effective isotropic consolidation
stress (σ'​c​1 in​ ​table 2).​ ​Specimens​ ​1, 2 and 3 were then sheared to failure under undrained
conditions without changing the isotropic consolidation stress, and the deviator stress and pore
pressure were recorded. Specimens 4 and 5 were unloaded to σ'​c ​2 before being sheared to failure
under undrained conditions. Figure 6 shows CU triaxial test results for the normally consolidated
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9. (NC) specimens (1, 2, 3) and figure 7 shows the results for the overconsolidated (OC) specimens
(4, 5).
Figure 6: ​Deviator Stress and pore pressure as a function of axial strain for NC specimens
during CU triaxial testing.
9

10.
Figure 7:​ Deviator stress and pore pressure as a function of axial strain for OC specimens during
CU triaxial testing.
10

11. Plotting Mohr circles at failure and Mohr­Coulomb failure envelope estimated from the
CU triaxial tests for the three normally consolidated specimen resulted in the following
parameters (figure 8):
ϕ’ = 33°
c’ = 0 psf
ᵰ​ff1​ = 1946.0 psf
ᵰ​ff2​ = 2363.3 psf
ᵰ​ff3​ = 2667.4 psf
FIGURE 8: ​Mohr circles at failure and the corresponding Mohr­Coulomb failure envelope
resulting from CU triaxial tests on NC specimens.
Plotting Mohr circles at failure and Mohr­Coulomb failure envelope estimated from the
CU triaxial tests for the two OC specimen resulted in the following parameters (figure 9):
ϕ’=31
c’ = 1000 psf
ᵰ​ff4​ = 1791.6 psf
ᵰ​ff5​ = 2170.1 psf
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12.
FIGURE 9: ​Mohr circles at failure and the corresponding Mohr­Coulomb failure envelope
resulting from CU triaxial tests on OC specimens.
The ratio of τ​ff​/σ’​c ​and the OCR for all 5 specimens is summarized in table 3.​ ​The
relationship between τ​ff​/σ’​c​ and OCR is shown in figure 10. Using the relationship between the
normalized shear strength and OCR, and the stress profile in the Quaternary terrace deposits
(using unit weights from samples taken in boring 3; Appendix 2), it is possible to calculate the
shear strength as a function of depth. The vertical effective stress profile assuming a deep
groundwater table (GWT), OCR profile, normalized shear strength profile, and the Strength
profile are shown in figure 11.
Table 3:​ OCR and normalized shear strength ratio​ ​resulting from CU triaxial tests for all 5
specimens.
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13.
Figure 10: ​Strength ratio vs. OCR of silty clay
Figure 11: ​Vertical effective stress, OCR, normalized shear strength, and shear strength as a
function of depth for the deep groundwater condition during drilling.
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14. Standard penetration testing (SPT)
SPTs were performed in borings 2, 3, 4 and 6 (appendix 2). The samples were retrieved
using a 63.5 kg hammer that was dropped 76 cm. The hammer was lifted by a cathead, with two
turns of the rope around the cathead. This method of release delivers about 60% energy
efficiency. The raw blow counts, N​60​, were recorded and then corrected for overburden resulting
in (N​1​)​60​ blow counts (equations 1­2). These results are presented in tables 4­7.
(N​1​)​60​ =C​N​N​60 ​(1)
C​N​=
√σvo′
Patm
(2)
Table 4: ​SPT results for boring 2
Table 5: ​SPT results for boring 3
Table 6:​ SPT results for boring 4
Table 7:​ SPT results for boring 6
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15. Figure 12 shows (N​1​)​60​ as a function of sampling depth for borings 2, 3, 4, and 6. Figure
13 shows (N​1​)​60​ as a function of depth, with lithology demarcated. Table 8 summarizes the
average strength parameters estimated using SPT for each soil type.
Figure 12: ​(N​1​)​60​ depth profile from combined SPT from borings 2, 3, 4, and 6.
Figure 13:​ SPT (N​1​)​60 ​with depth, divided by soil type.
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16.
Table 8: ​Summary of the average relative density and strength parameter evaluated using SPT
for each soil type.
Compaction test
A Modified Proctor Compaction Test (ASTM D1557) was performed on a specimen
taken from the bulk sample of soil from the borrow site (Appendix 3). Five specimens were
moisture conditioned to different water contents, and compacted with the specified energy level
into the standard mold. The specific gravity of the solids in this soil is Gs = 2.70. The results are
indicated below in table 9.
w (%) 4.5 6.4 8.1 11.0 12.1
γ​d​ (lb/ft​3​
) 121.5 130.0 133.0 127.1 124.0
Table 9: ​Modified Proctor compaction test results.
The dry densities for different water contents on the zero air voids (ZAV) line have been
summarized in​ ​table 10.
Table 10: ​Water contents and dry densities on the ZAV.
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17. The Modified Proctor compaction test (ASTM D1557) gives an optimum moisture
content of 8.0% and a maximum dry density of 133.f lb/ft​3​
(figure 14).
Figure 14: ​Modified Proctor compaction curve for fill.
Hydro­compression tests were performed after soaking the soil and applying surcharges
of 900 pounds per square foot (psf), 1800 psf and 2700 psf. The test results are shown in table 11
, and the contour lines on the compaction curves are displayed in figure 15 (see also appendix 3
for ASTM D1883 results).
Table 11: ​Hydro­compression test results.
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18.
Figure 15: ​Volumetric strain of fill after soaking under (a) 900 psf, (b) 1800 psf, and (c) 3600
psf surcharges.
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19. Direct shear Test
Based on our recommendation, test fills were placed by a contractor and then
subsequently we performed direct shear tests on samples from the compacted fill. The fills were
placed in two areas, and we performed borings in two locations (boring 601 and boring 602)
using a 24­inch diameter bucket auger. Since the samples had sufficient binder material (silt and
clay) such that relatively undisturbed samples could be obtained, we obtained Shelby Tube
samples from borings 601 and 602 at depths of 3 feet. We then ran direct shear tests (ASTM
D3080) on the two samples (figure 16).
Figure 16​: Direct shear test results from samples from boring 601 and 602.
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20. ANALYSIS AND RECOMMENDATIONS
Grading and Fill Placement
The location and thickness of the proposed artificial fill is shown in the mass grading
plan (appendix 1). The fill will be borrowed on­source; in other words the fill will be excavated
from the site. In order to prevent failure due to differential movement of structural components,
it is necessary that the volumetric strains of the fill under surcharge pressure after soaking are
limited to 0.5% or less for swell and 1.0% or less for hydro­compression.
Compaction tests performed on samples of the borrowed fill are shown in figures 14 and
15. The shaded areas in figure 17 represent the dry unit weight and water contents that meet the
design criteria of limiting swell upon wetting to less than 0.5% and hydro­compression to less
than 1.0%.
It is recommended that for fills less than 13 feet in depth (expected to have overburden
pressures of 900­1800 psf), the minimum relative compaction should be 89% of the maximum
dry density obtainable according to the ASTM D1557 compaction test and a water content
between optimum to 2% over optimum. For fills greater than or equal to 13 feet (expected to
have overburden pressures greater than 1800 psf), the relative compaction should be greater than
94% of the maximum dry density according to the ASTM D1557 compaction test, and a water
content between optimum and 4% over optimum.
20

21.
Figure 17​: Design criteria (shown in shaded regions) for volumetric strain of fill after soaking
and applied surcharges of (a) 900 psf, (b) 1800 psf, and (c) 3600 psf
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22. Slope Stability During Site Excavation
Although the Occupational Safety and Health Administration (OSHA) provides
requirements for maximum slope heights, inclinations, and configurations that are allowable for
temporary slopes, it is possible that the OSHA prescriptive guidelines can be superseded by a
site­specific design provided by a registered professional engineer. It was asked if our firm could
analyze the stability of temporary 1:1 slopes to be used during the excavation of the site. The
criteria used by our firm is that any temporary cut that has a slope must have a factor of safety
against failure of 1.5.
As a conservative estimate, the longest possible 1:1 slope possible during excavation
given the existing topography and the mass grading plan was identified and used for slope
stability analysis (appendix 1). Additionally, the entire slope and toe material upon which it sits
is assumed to be a uniform, frictionless clay.​ ​The groundwater table was assumed to be deep
(below the potential sliding surfaces), and conditions were taken throughout the slide mass as
effectively saturated due to precipitation. The saturated unit weight (γ​s​) of the clay used in the
analysis, 138.0 pcf, was calculated from the measured dry unit weight of a clayey terrace deposit
sample taken at a depth of 17 feet from boring 3 (appendix 2). The undrained shear strength (S​u​)
profile​ ​was​ ​calculated by de­normalizing the normalized shear strength profile (figures 10, 11).
The clay slope was sub­layered such that within each sublayer S​u​ can be approximated as
constant​ ​(figure 18).
Figure 18: ​Schematic of the slope used in our stability analysis, with the undrained shear
strength as a function of depth for each clay sublayer shown on the right.
A Swedish Circle analysis was performed for a variety of geometries in order to find the
critical geometric constraint (Duncan et al., 2014). Using the identified critical geometry, a
global search using the software package Rocscience was used to find the critical circle center
location. Additionally, a translational wedge failure was considered (Duncan et al., 2014). Of
those tested, the critical failure surface was a toe circle, with a the lowest global minimum factor
of safety of 1.32 found by iterating over circle center locations (figure 19).
22

23.
Figure 19:​ Screenshot from the Rocscience software package showing the failure surface with
the lowest global factor of safety.
We recommend against the use of a 1:1 temporary slope, as the global minimum factor of
safety is less than the allowable 1.5. Instead, we recommend that the slope be flatter than 1:1 and
that there be a three­foot horizontal bench if the vertical height of the slope exceeds 20 feet.
Moreover, slopes that will be in place for more than a few days should have the ground surface
above the top of the slope inclined away from the slope to prevent erosion of the slope during
potential rain storms. Further analysis is necessary for more specific recommendations.
Retaining Walls
Free­standing retaining walls will be located at different locations on the hotel site
(outside of the footprint of the main hotel building), with a height of up to 15 feet. The retaining
walls must be designed to resist lateral loads caused by a surcharge pressure of 300 psf. The
free­standing retaining walls will be built on flat land and must have a minimum sliding factor of
safety of 1.5 and a minimum bearing capacity factor of safety of 2, in accordance with ASCE
1994 design standards.
The backfill will be placed behind the retaining walls after construction and will consist
of the on­site SM­SC fill material. The unit weight of the backfill (γ​backfill​) used in the analysis,
131 pcf, was calculated using dry densities and optimum water content from the Modified
Proctor Compaction Test (ASTM D1557). The lower two feet of the fill behind retaining wall
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24. will be compacted to at least 94% of the maximum dry density obtainable with the ASTM D1557
method of compaction, and the upper 13 feet will be compacted to at least 89% of the maximum
dry density obtainable with the ASTM D1557 method of compaction. Corresponding to this, the
water content will be compacted to in the range of optimum water content to 4% over optimum.
Direct shear tests (ASTM D3080) were performed on fill samples from Borings 601 and
602 after two test sections of fill were placed. From the two samples, an average friction angle of
25° was derived. The foundation soil of is a clean sand layer of terrace deposits with a unit
weight (γ​field​) of 105 pcf and a friction angle of 36°, from SPT (table 8). All values are calculated
assuming good drainage.
A method of Rankine earth pressure analysis was utilized to predict active and passive
earth pressures acting on the free­standing retaining walls. It can be assumed that the soil is
cohesionless and the wall is frictionless, as outlined in Coduto et al 2010. The freestanding
retaining walls are analyzed using conventional stability analysis (ASCE 1994) to determine the
wall’s ability to withstand sliding, overturning and bearing capacity. The retaining walls have an
active earth pressure of 5452 lb/ft resulting from the backfill. Additionally, the surcharge
pressure present on the ground surface behind the wall will create an active earth pressure of
1800 lb/ft. The walls will also have a passive pressure of 1450 lb/ft. Figure 20 illustrates the
variation of earth pressure with depth. For the free­standing cantilever retaining walls, a footing
width of 15.5 ft is required to satisfy the aforementioned design conditions.
Figure 20: ​Lateral Earth Pressures vs. Depth for Retaining Walls
It is recommended to use a 15 ft high cantilever concrete retaining wall. We recommend
using a shallow spread footing foundation with a minimum width of 15.5 feet. These footing
dimensions will ensure a sliding factor of safety of at least 1.5 and a bearing capacity factor of
safety well over 2, as required by the ASCE 1994 design criteria.
The above pressures assume that sufficient drainage will be provided behind the walls to
prevent the build­up of hydrostatic pressures from surface and subsurface water infiltration.
Adequate drainage may be provided by a subdrain system consisting of a 4­inch rigid perforated
pipe bedded in ¾­inch clean, open graded rock. The entire rock/pipe unit should be wrapped in
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25. an approved, non­woven polyester geotextile. The rock and fabric placed behind the wall should
be at least one foot in width and should extend to within one foot of finished grade. The upper
one foot of backfill should consist of onsite, compacted soil of low permeability. The subdrain
pipe should be connected to a system of closed pipes that lead to suitable discharge facilities.
These ditches, which will collect runoff water from the slopes, should be sloped to drain
to suitable discharge facilities. The top of the walls should extend at least one foot above the
ditch. All structural backfill placed behind the walls should be compacted to at least 90 percent
relative compaction based on the ASTM D1557 standard.
Basement Walls
The proposed hotel will have basement walls due to the stepping nature of the basement
up the existing grade. These walls will have a maximum height of 20 feet. It is assumed that the
basement wall will move negligibly when the backfill is being placed. The basement walls will
be built on flat land and must have a minimum sliding factor of safety of 1.5 and a minimum
bearing capacity factor of safety of 2, in accordance with ASCE 1994 design standards.
The backfill material that will be placed behind the basement walls after construction will
be similar to the backfill placed behind the retaining walls. It will consist of the on­site SM­SC
material with a unit weight (γ​backfill​) of 131 pcf. It will have an average friction angle of 25 °, as
derived from the direct shear tests (ASTM D3080, figure 16). The foundation soil of is a clean
sand layer of terrace deposits with a unit weight (γ​field​) of 105 pcf and a friction angle of 36°,
from SPT (table 8). All values are calculated assuming good drainage.
The basement walls do not need to undergo stability analysis as there little potential for
stability failure when the backfill is placed. The basement walls have a coefficient of earth
pressure of 0.57 and and an at rest earth pressure of 14934 lb/ft. Figure 21 illustrates the
variation of earth pressure with depth for the basement walls. Sufficient drainage should be
provided behind the basement wall using the same methodology as for the free­standing
retaining walls.
Figure 21: ​Lateral Earth Pressures vs. Depth for Basement Walls
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26.
Hotel Foundations
The following foundation design conditions were taken from either the RFP, or from
local / state / federal building codes, or from generally accepted geotechnical engineering
practices:
● The hotel foundation columns will be each spaced 30 ft. apart.
● The maximum differential settlement between the Type 1 footings must be less than ½
inch, and less than 1 inch for between Type 3 footings.
● The indicated loads for each footing type, based on sub­footing material and based on
location:
○ Type 1: 350­700 kips (clay, sand & bedrock), located along C­C’
○ Type 2: 200­350 kips (bedrock), located along B­B’
○ Type 3: 250­450 kips (sand), located along B­B’
● The factor of safety must be greater than 3.0 against bearing cap failure for each footing.
● For each footing type, two embedment depths will be examined: 2 & 5 ft., to evaluate
strength gains provided by deeper footings.
Additionally, all the previous testing designations and specifications given up to this
point are assumed to have been met. Any other assumptions of trivial nature that were made in
the following analysis are presented in their respective sections of this report.
The properties of the soil are as follows:
Type 1 Footings (Clay, Sand & Bedrock):
The footings for this section of the foundation will be arranged adjacently on bedrock
(San Onofre Breccia) and soil (Marine Terrace Deposits). The bedrock is assumed to be
incompressible and to have excess bearing strength relative to the soils. The soils present at this
location are layered silty sand and clayey sand (referred in this report as ‘clay’), sand, and
bedrock (figure 22). The groundwater table, initially beneath the bedrock, will be assumed to rise
to the base of the foundation due to irrigation, which will create saturated soil conditions. The
rise of the groundwater table will increase the overconsolidation ratios for all soil sublayers that
were previously above the initial groundwater table. Additionally, this means that the undrained
shear strengths will be used for the clay analysis (figure 23). For foundations located on soils, the
layering will be assumed to be clay on top, with a sandy layer underneath, and bedrock beneath
that. The sand is assumed to have immediate settlement, while the clay is presumed to
consolidate over time. The column load for this section, as stated in the design conditions, is 700
kips.
26

27.
Figure 22​: Structure of sub­footing soil for Type 1 footings, prior to the implementation of the
mass grading plan. However, little to no grading is expected to occur at this location. This was
determined from Boring 3 performed by Profit, Inc. (appendix 2). The perforated lines indicate
the sublayer boundaries used in the analysis.
Table 12:​ Property values for each clay sublayer in Figure 22, (top to bottom), as evaluated
during laboratory testing (appendix 3).
27

28.
Figure 23: ​Undrained shear strength profile for clay beneath Type 1 footings. The undrained
strengths were used, because the groundwater table is assumed to rise due to irrigation around
the hotel.
Type 2 Footings (Bedrock):
The Type 2 footings will rest entirely on solid bedrock, which is assumed to have
significantly higher strength parameters than the surrounding soils. As such it will experience
effectively zero settlement. The columns loading for this section is expected to be 350 kips.
Type 3 Footings (Sand):
The Type 3 footings will be arranged on sandy deposits with bedrock underneath (figure
24). The sand is considered to experience immediate settlement, and to drain rapidly and thus not
cause any appreciable increase in pore pressure. The following charts outline the relevant
properties for this soil (Table 13). The columns loading for this segment were indicated as 450
kips.
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29.
Figure 24:​ Structure of sub­footing soil for Type 3 footings, prior to the mass grading plan
implementation. Determined from Boring B­13 performed by Moore and Taber (appendix 2).
Table 13:​ Relevant initial properties of sand beneath Type 3 footings. The unit weight for the
sands was assumed relatively uniform, and calculated to be 109 pcf in the laboratory testing
(appendix 3).
The analysis was carried out in the order of settlement first, then bearing capacity second
in order to evaluate which would be the controlling factor in the footing design. For the clay
settlement calculations (Type 1 Footings), the Terzaghi 1­D model (Holtz et al., 2011) was used,
using the double drainage variable due to the presence of the sand layer beneath the clay. The
clay was divided into four representative sub­layers (figure 22), then analyzed layer by layer.
The sand settlements (Type 3 Footings) were calculated using the Burland and Burbidge 1985
procedure. Finally, the Meyerhof general bearing capacity equation was then used in each case to
evaluate the maximum allowable bearing pressure for the foundations (Das, 2010).
After retrieving the laboratory test results and running the analysis procedures outlined
above, the following results were obtained. For Type 1 footings, the bearing capacity is the
controlling factor of foundation dimensions, for Type 2 footings, the bearing capacity and
settlement are considered irrelevant, and for Type 3 footings, the controlling factor is the
settlement. Three footing dimensions were tested for each case, and two embedment depths were
tested. For the final design for Type 1 footings, the expected settlement on the clay is 0.95
inches, and it is expected to occur over time, with 43% (0.41 inches) settlement after 5 years
(figures 25 and 26). For the Type 2 footings, there is no expected settlement, because it is
situated directly on the bedrock. For the final design for Type 3 footings, the sand is expected to
settle 0.96 inches almost immediately.
29

30.
Figure 25: ​Percent settlement over time for Type 1 footings on clay soil
Figure 26: ​Total settlement in inches for Type 1 footings on clay soil.
After assessing the results, our recommended design specifications for the hotel building
footings are as follows:
For the Type 1 footings, square spread footings should be used with a side length of 19
feet and at an embedment depth of a minimum 5 feet. This gives an allowable bearing capacity
of 2,500 psf, a factor of safety greater than 3, and meets the previously outlined design criteria.
Type 2 footings can be modeled after either Type 1 or Type 3, due to the lack of
settlement expected from the bedrock, so long as they extend entirely down to the bedrock.
Further testing and analysis would be required to design a separate, bedrock­specific footing.
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31. Type 3 footings should be square shallow spread footings, with 13 feet per side and
embedded at a minimum of two feet below the surface, which will give a bearing capacity of
2,500 psf. This gives a factor of safety greater than 3 and meets the previously outlined design
criteria.
Slope Stability of Coastal Bluffs
The sea bluffs on the proposed hotel site also present a potential landslide threat due to
the steep nature of the slope; it is of interest to minimize the instabilities in that location. Based
on historic observations that periodic stability failures of the fill along the bluff occur after
rainstorms, we back­calculated the minimum shear strengths for the fill material (Duncan et al.,
2014). It is assumed that the water table depth is below the potential sliding surface, and there is
an effectively saturated condition throughout the slide mass due to precipitation. The factor of
safety is assumed to be 1.0 when the fill is in a saturated condition. A wedge passing through the
fill material at the edge of cross­section B­B’ is used, with the thickest fill at the edge of the bluff
(figure 27). Using a lower­bound effective stress friction angle of 30 degrees, the effective stress
cohesion of the fill during a slope failure is estimated to be 14.68 psf. Our recommendations for
the area of the site along the bluff include installing a fence along the bluff in order to keep hotel
occupants away from the zone of potential failure, and to install a subdrain behind the slope to
increase drainage in the area. Additionally we recommend that the area behind the bluff be
graded such that it slopes away from the bluff, in order prevent flow of water and soil materials
down the bluffs.
Figure 27:​ Assumed failure geometry along sea bluffs used to estimate the minimum necessary
effective stress cohesion of artificial fill.
Restaurant Foundation
The restaurant building will be constructed close to the edge of the bluff, as shown in
figure 28. Because of the restaurant’s proximity to the bluff and anticipated gradual erosion of
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32. the bluff, the best option is to use deep (drilled shaft) foundations to support the restaurant so that
the stability won’t be compromised if erosion ever gets to the edge of the restaurant building.
The sum of dead and live loads from the restaurant structure (averaged over the full footprint of
the building), including floor slab loading, is 250 psf. The foundations for the restaurant structure
must be adequate to support these loads, and in this case, because of the proximity of the bluff
and the potential erosion that might occur beneath the building, the floor slab must also be
structurally supported. The factor of safety of the deep foundations must be at least 2 for side
load transfer and 3 for end bearings.
Figure 28: ​End of Cross Section C­C
The drilled shafts will be 2 feet below grade in a layer of sandy Terrace deposits followed
by a bedrock layer composed of San Onofre Breccia. The shafts will extend 5 feet into the
Terrace deposits at its shortest depth to rock and 7 feet at its longest depth to rock. Below the
terrace deposits, the shafts will extend 5 feet into the bedrock. The unit weight (γ) of the sandy
Terrace deposits is 109 pcf and the friction angle (φ′ ) of the sandy Terrace deposits is 36.8°. The
unit weight (γ) of the breccia is 135 pcf and its unconfined compressive strength is 10,000 psf.
The bedrock has a rock quality designation (RQD) of 1.
The restaurant is in close proximity to the bluffs, which are anticipated to gradually
erode. Because of the weakness of the soil underneath the restaurant, using a drilled shaft
foundation is more advisable than shallow foundations. The drilled shafts can safely transfer the
loads of the restaurant into more competent layers of soil and rock below. The axial loads were
assessed using static methods of analysis. The soil was found to have a lateral earth coefficient
(K) of 0.403 and a friction angle (δ) of 36.6°. The effective stress for the shortest shaft and
longest shaft are computed according to the sublayers of the sandy Terrace deposits and is
tabulated in table 14 and table 15, respectively.
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33.
Depth to the top of layer (ft) Depth to the bottom of layer
(ft)
σ​h​’ (psf)
(middle of sublayer)
0 1 97.24
1 2 116.02
2 3 134.80
3 4 153.58
4 5 172.36
Table 14: Effective Stress Values for the Shortest Drilled Shaft
Depth to the top of layer (ft) Depth to the bottom of layer
(ft)
σ​h​’ (psf)
(middle of sublayer)
0 2 106.63
2 4 144.19
4 6 181.75
6 7 235.07
Table 15: Effective Stress Values for the Longest Drilled Shaft
The bedrock is assumed to be cohesive intermediate geomaterial (IGM), with an adhesion
factor of 0.2, obtained from FHWA 2010. The bedrock has a slide resistance of 2000 psf and a
tip resistance of 25000 psf. Group effects may be neglected for the restaurant’s drilled shafts.
The shafts will be 18 inch diameter, cast­in­place reinforced concrete, and installed using
the dry method. The capacity of the shaft varies with its length. The shortest drilled shaft will
have a total allowable capacity of 39.46 kips and the longest drilled shaft will have a total
allowable capacity of 40.2 kips. Based on these capacities, the drilled shafts may be spaced with
approximately 12 feet between each shaft in order to withstand the 250 psf load from the
restaurant. Additionally, we recommend that the floor slab be reinforced with pre­stressed steel
girders to ensure the structural integrity of the floor slab.
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34. Plan Review and Construction Observation
In ensuring the safety and soundness of the analysis presented, we recommend that a
suitably qualified geotechnical engineer review the project plan and specifications before you
proceed with construction in order to check conformance with the recommendations put forward
in this report. Additionally, we recommend that a qualified geotechnical engineer have the
opportunity to observe the construction and ensure that everything is completed in accordance
with the recommendations in this report. Lastly, we recommend an in­depth five­year evaluation
of the site and structures to ensure it is performing as expected. We will not claim any liability if
the recommendations set forth in this report are not implemented during the subsequent phases of
this project.
If you have any questions concerning this matter, or we can be of any further service to
you, please contact us.
Yours sincerely,
Kamal Deep Sheokand
Principal Engineer
Grace A. Parker
Project Engineer
Adam C. Richardson
Staff Engineer
Pranasha Shrestha
Project Engineer
34

35. REFERENCES
Burland, JB, and Burbridge, MC (1985). ‘‘Settlement of foundations on sand and gravel.’’
Proc. Inst. Civ. Eng., Struct. Build., ​78(6), 1325– 1381.
Coduto, DP, MR Yeung, and WA Kitch (2010). ​Geotechnical Engineering: Principles &
Practices​, 2​nd​
ed., Prentice Hall.
Das, BM (2010). ​Principles of Foundation Engineering​, 7​th​
ed., Cengage Learning ­
Engineering.
Duncan, JM, SG Wright, and TL Brandon (2014). ​Soil Strength and Slope Stability​, 2​nd​
ed.,
Wiley ­ Technology and Engineering.
Holtz, RD, WD Kovacs, and TC Sheahan (2011). ​An​ ​Introduction to Geotechnical Engineering​,
2​nd​
ed., Pearson ­ Engineering Geology.
“Retaining and Flood Walls.” (1994) ​Technical Engineering and Design Guides as adapted from
the US Army Corps of Engineers, ​No. 4, ASCE Press.
ATTACHMENTS: ​appendix 1, appendix 2, appendix 3.
35

36. APPENDIX 1: ​Plot plan, geologic cross sections A­A’, B­B’, and C­C’, mass grading plan.
APPENDIX 2: ​Boring logs from prior investigations (Moore and Taber, 1980, and AGRA Earth
and Environmental, Inc., 1999), and from our investigation.
APPENDIX 3​: Additional lab tests (sieve analysis,​ ​gradation and Atterberg limits, compaction
CBR and swell tests (ASTM D 1883))
36